Mj_mar.-apr._2016_v._113_no._2 .pdf

V. 113, NO. 2 MARCH-APRIL 2016

ACI MATERIALS

J O U R N A L

A JOURNAL OF THE AMERICAN CONCRETE INSTITUTE

CONTENTS Board of Direction President Sharon L. Wood Vice Presidents Michael J. Schneider Khaled Awad Directors Dean A. Browning JoAnn P. Browning Cesar A. Constantino Alejandro Durán-Herrera Augusto H. Holmberg Kimberly Kayler Cary S. Kopczynski Kevin A. MacDonald Fred Meyer Michael M. Sprinkel Roberto Stark David M. Suchorski Past President Board Members William E. Rushing Jr. Anne M. Ellis James K. Wight Executive Vice President Ron Burg

Technical Activities Committee Trey Hamilton III, Chair Matthew R. Senecal, Secretary Michael C. Brown JoAnn P. Browning Catherine E. French Fred R. Goodwin Larry Kahn Neven Krstulovic-Opara Kimberly Kurtis Tracy D. Marcotte Jan Olek Michael Stenko Andrew W. Taylor Eldon G. Tipping

Staff

Executive Vice President Ron Burg Engineering Managing Director Michael L. Tholen

ACI Materials Journal March-April 2016, V. 113, No. 2 a journal of the american concrete institute an international technical society

131 Influence of Fiber Orientation on Bridging Performance of Polyvinyl Alcohol Fiber-Reinforced Cementitious Composite, by Toshiyuki Kanakubo, Masaru Miyaguchi, and Kohei Asano 143 Strain Rate Sensitivity of Fiber-Reinforced Cementitious Composites, by H. Othman and H. Marzouk 151 Analysis of Compressive Strength Development and Carbonation Depth of High-Volume Fly Ash Cement Pastes, by Xiao-Yong Wang and Ki-Bong Park 163 Behavior of Anchored Carbon Fiber-Reinforced Polymer Strips Used for Strengthening Concrete Structures, by Wei Sun, James O. Jirsa, and Wassim M. Ghannoum 173  Effect of Dosage of Fly Ash and NaOH on Properties of Pisha Sandstone-Based Mortar, by Changming Li, Tingting Zhang, and Lijiu Wang 185 Compressive and Time-Dependent Strength of Concrete Masonry Constructed with Type M Mortar and Grouts Containing High Volume of Fly Ash and Slag, by Fernando S. Fonseca, Scott M. Watterson, and Kurt Siggard 197 Compatible Datum Temperature and Activation Energy for Concrete Maturity, by Chang Hoon Lee and Kenneth C. Hover 207 Performance of Full-Scale Self-Consolidating Rubberized Concrete Beams in Flexure, by Mohamed K. Ismail and Assem A. A. Hassan 219  Tensile Behavior of Steel-Polypropylene Hybrid Fiber-Reinforced Concrete, by Lihua Xu, Le Huang, Yin Chi, and Guodong Mei 231 Nano-Modified Fly Ash Concrete: A Repair Option for Concrete Pavements, by A. Ghazy, M. T. Bassuoni, and A. Shalaby 243

Reviewers in 2015

Managing Editor Jerzy Z. Zemajtis Staff Engineers Khaled Nahlawi Matthew R. Senecal Gregory M. Zeisler Publishing Services Manager Barry M. Bergin Editors Carl R. Bischof Tiesha Elam Kaitlyn J. Hinman Kelli R. Slayden Editorial Assistant Angela R. Matthews

Discussion is welcomed for all materials published in this issue and will appear ten months from this journal’s date if the discussion is received within four months of the paper’s print publication. Discussion of material received after specified dates will be considered individually for publication or private response. ACI Standards published in ACI Journals for public comment have discussion due dates printed with the Standard. ACI Materials Journal Copyright © 2016 American Concrete Institute. Printed in the United States of America. The ACI Materials Journal (ISSN 0889-325x) is published bimonthly by the American Concrete Institute. Publication office: 38800 Country Club Drive, Farmington Hills, MI 48331. Periodicals postage paid at Farmington, MI, and at additional mailing offices. Subscription rates: $166 per year (U.S. and possessions), $175 (elsewhere), payable in advance. POSTMASTER: Send address changes to: ACI Materials Journal, 38800 Country Club Drive, Farmington Hills, MI 48331. Canadian GST: R 1226213149. Direct correspondence to 38800 Country Club Drive, Farmington Hills, MI 48331. Telephone: +1.248.848.3700. Website: http://www.concrete.org.

ACI Materials Journal/March-April 2016

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Contributions to ACI Materials Journal

MEETINGS APRIL 10-12—NRMCA’s Annual Convention, San Diego, CA, www. nrmca.org/Conferences_Events/ AnnualConvention/2016/index.html 10-13—GeoAmericas 2016, Miami Beach, FL, www.geoamericas2016.org 20-23—10th Erbil International BuildingConstruction, Municipality Equipment, Machinery & Natural Stone Exhibition, Erbil, Iraq, http://erbilbuilding.com/index. php/visitors/2014-04-10-09-56-16 24-26—2016 PTI Convention, Long Beach, CA, www.post-tensioning.org/page/ PTI-Convention 27-29—The 6th Amazon & Pacific Green Materials Congress and Sustainable Construction Materials LAT-RILEM Conference, Cali, Colombia, www.6gmc. com.co/paginas/welcome

MAY 3-4—Missouri Concrete Conference, Rolla, MO, www.concrete.mst.edu 10-12—SDC Technology Forum #39, San Antonio, TX, www.concretesdc.org 15-18—International Concrete Sustainability Conference, Washington, DC, www.scc2016.com

23-25—Concrete Service Life Extension Conference, Orlando, FL, http://concrete. nace.org 24-25—11th Global Slag Conference, London, UK, www.globalslag.com/ conferences/global-slag/introduction 25-26—5th Annual Modular & Precast Construction, Bangkok, Thailand, www. trueventus.com/event.php?intid=316 26-29—Construction History Society of America 5th Biennial Meeting, Austin, TX, www.chsa-5thbiennial.org

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JUNE 7-9—8th RILEM International Conference on Mechanisms of Cracking and Debonding in Pavements, Nantes, France, http://mcd2016.sciencesconf.org 7-9—Knowledge Exchange for Young Scientists (KEYS): Sustainable Cement and Concrete Construction – Improvement of Solid Waste Management, Accra, Ghana, www.rilem. org/gene/main.php?base=600040#next_929

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THE ACI CONCRETE CONVENTION AND EXPOSITION: FUTURE DATES 2016—April 17-21, Hyatt & Wisconsin Center, Milwaukee, WI 2016—October 23-27, Marriott Philadelphia, Philadelphia, PA 2017—March 26-30, Detroit Marriott at the Renaissance Center, Detroit, MI 2017—October 15-19, Disneyland Hotel, Anaheim, CA For additional information, contact: Event Services, ACI 38800 Country Club Drive Farmington Hills, MI 48331 Telephone: +1.248.848.3795 e-mail: [email protected]

ON COVER: 113-M13, p. 136, Fig. 12—Tensile test specimen. (Note: 1 mm = 0.0394 in.)

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ACI Materials Journal/March-April 2016

ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M13

Influence of Fiber Orientation on Bridging Performance of Polyvinyl Alcohol Fiber-Reinforced Cementitious Composite by Toshiyuki Kanakubo, Masaru Miyaguchi, and Kohei Asano Crack bridging performance of fibers strongly affects the tensile characteristics of fiber-reinforced cementitious composites (FRCCs) after first cracking. The fiber orientation distribution is likely to be affected by factors that include fresh-state properties, casting method, formwork geometry, and others. The objective of this study is to investigate the influence of the fiber orientation on the bridging performance in polyvinyl alcohol (PVA) FRCCs through a visualization simulation using a water glass solution and a calculation of the bridging law. The main parameter of the investigations in the present study is the casting direction. To evaluate the fiber orientation distribution quantitatively, an approximation methodology using an elliptic function is newly introduced. The bridging stress versus crack width relationship is calculated considering the elliptic distribution, the snubbing effect, and the fiber strength degradation. The calculated stress-crack width curves can express the uniaxial tension test results after first cracking well. Keywords: bridging law; casting direction; elliptic function; fiber orientation; fiber-reinforced cementitious composites; image analysis; orientation intensity; uniaxial tension test.

INTRODUCTION The crack bridging performance of fibers, which is generally expressed by a bridging stress-crack opening relationship (called the bridging law), strongly affects the tensile characteristics of fiber-reinforced cementitious composites (FRCCs) after first cracking. The bridging performance is characterized and/or controlled by the properties of the matrix, the fiber, and the fiber-matrix interface.1,2 Since the 1980s, studies on high-performance fiber-reinforced cement composites (HPFRCCs) and engineered cementitious composites (ECCs) have been conducted to understand the crack bridging performance, primarily because these composites require the balanced properties of the matrix, the fiber and their interface, to exhibit the pseudo strain-hardening behavior.3,4 One of the examples of a polymeric fiber bridging law is that presented by Kanda and Li,5 who described it for polyvinyl alcohol (PVA) fibers assuming the following characteristics: 1) the chemical bond in the fiber-matrix interface; 2) the rupture of the fiber; and 3) the tensile strength reduction owing to inclined-angle bridging. These considerations had primarily been introduced to account for the characteristics of randomly oriented, discontinuous fibers.6 Many researchers have studied the effects of fiber orientation on the mechanical characteristics of FRCC, including fiber-reinforced concrete (FRC). The categories of these materials including HPFRCC and ECC are summarized in some literatures.7 In addition, self-consolidating concrete (SCC) and ultra-high-performance fiber-reinforced concrete (UHP-FRC) have been developed for the last several ACI Materials Journal/March-April 2016

decades. These materials have specific properties that require researchers and engineers to be attentive to fiber orientation. The scheme of the current approach to evaluate the fiber orientation has considered the casting method, fresh-state properties, flow, vibration, and formwork geometry.8 The cementitious matrix used in HPFRCC and ECC has a high viscosity, aiding the random distribution of the fine fibers and commonly has self-consolidating properties. These characteristics indicate that the bridging law in HPFRCC and ECC is likely to be affected by the fiber orientation. In fact, the tensile characteristics of polymeric FRCCs differ because of the casting direction and the dimension of the specimen.9 The wall effect, in which the fiber orientation is influenced by the surface of the mold, has also been studied by many researchers. Li and Wang10 categorized the fiber orientation as two-dimensional (2-D) random and three‑ dimensional (3-D) random by the specimen dimensions in two perpendicular sectional planes. The ultimate tensile strain of PVA-ECC tends to decrease if the specimen dimension changes from 2-D to 3-D. Statistical approaches on the fiber orientation distribution began in the 1960s. Naaman11 proposed a sinusoidal function as the probability density function (PDF) of the angle between the fiber and the normal vector of the cut plane. Stroeven12 indicated the combination of three typical distributions—namely, 3-D random, 2-D random, and perfectly aligned one-dimensional—for simulation of arbitrary orientation distributions. One of the examples of the approaches adopted to study the wall effect is presented by Dupont and Vandewalle.13 They proposed a theoretical quantification by predicting the total number of fibers crossing a rectangular section. Considering the influence of the matrix flow, Xia and Mackie14 proposed the probabilistic spatial orientation using the beta distribution as the axisymmetric fiber orientation. There have been many studies to investigate the fiber orientation and distribution by experimental approaches observing fibers directly via image-based analysis. In case of steel fibers, the X-ray technique is one of the effective methods. Recently, microcomputed tomography (micro-CT) has been used to characterize the fiber distribution.15 In case of polymeric fibers such as PVA, image analysis taking advantage of absorbing the ultraviolet radiation was conducted.16 ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2014-369.R2, doi: 10.14359/51688633, received June 15, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

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Table 1—Mechanical properties of PVA fiber and target fiber Type

Density, g/cm3 (lb/ft3)

Length, mm (in.)

Diameter, mm (in.)

Tensile strength, N/mm2 (ksi)

Elastic modulus, kN/mm2 (ksi)

PVA

1.30 (81.2)

12 (0.47)

0.10 (3.9 × 10 )

1200 (174)

28 (4060)

Nylon

1.14 (71.2)

12 (0.47)

0.24 (9.3 × 10 )

65 (9.4)

—

–3 –3

Table 2—Mixture proportion of HPFRCC Fiber volume Waterfraction, % binder ratio 2.0

0.39

Sandbinder ratio 0.50

Unit weight, kg/m3 (lb/yd3) Water

Cement

Fly ash

Sand

380 (641)

678 (1144)

291 (491)

484 (817)

Notes: Cement is high-early-strength portland cement; fly ash is Type II of Japanese Industrial Standard (JIS A 6202); sand is size under 0.2 mm (7.9 × 10–3 in.); high-range water-reducing admixture is binder × 0.6%.

It is considered that the bridging performance—that is, the tensile properties of FRCC—can be characterized using the fiber orientation distribution. The main objective of this study is to investigate the influence of fiber orientation distribution on the bridging law of polymeric fibers. The main experimental parameter selected in this study is the casting direction, which is considered to have an influence to the fiber orientation distribution. To achieve the goal, a visualization simulation is conducted using sodium silicate solution (known as water glass) to observe the flow patterns of the fibers in the tension test specimen. The results of the visualization simulation are discussed mainly for the distribution of the angles of the fibers. In this study, based on the visualization results, a new PDF is proposed to describe variation in the fiber angle. The PDF is expressed by two parameters: the principal orientation angle and the orientation intensity. These parameters indicate the angle and the tendency of the fibers to orient along the direction of the principal orientation. Finally, the bridging law, which is obtained by a numerical calculation, is compared with the tension test results. The fine fibers with a diameter ranging between 0.01 to 0.04 mm (4 × 10–4 to 16 × 10–4 in.) are commonly used for HPFRCC/ECC to actualize the pseudo-strain-hardening behavior and multiple cracking. On the other hand, multiple cracking makes the observation of the bridging law difficult. In this study, PVA fiber with a diameter of 0.10 mm (3.9 × 10–3 in.) is used to observe the bridging law (tensile stresscrack width curve) directly by the tension test subjected to single crack formation. RESEARCH SIGNIFICANCE Evaluation of the bridging law, accounting for the fiber orientation distribution, is necessary for predicting the precise tensile characteristics of FRCC. The fiber orientation, which is affected by casting method, fresh-state properties, flow, and formwork geometry, should be considered in the manufacturing of the composites for practical uses. The bridging characteristics of polymeric fibers are affected by their angle with the cracking plane. Understanding the fiber behavior expressed by the bridging law can facilitate understanding the tensile characteristics of FRCC. A simple mathe-

132

Fig. 1—Flowability test using funnel. (Note: 1 mm = 0.0394 in.) matical expression for the PDF of the fiber orientation distribution would also simplify simulations of the bridging law. VISUALIZATION SIMULATION OF FIBER ORIENTATION Materials for simulation test PVA fibers 0.10 mm (3.9 × 10–3 in.) in diameter were used in this study. The mechanical properties of PVA fibers are listed in Table 1. To visualize the flow of the fiber in a matrix, a sodium silicate solution (hereafter referred to as water glass) was adopted as the matrix. Water glass has high viscosity, and it is colorless and transparent. In regards to the practical use of ECC, the rheology of mortar matrix before mixing the fiber was inspected using the flow time,17 based on “Test method for flowability of grout for prestressing tendons (JSCE-F531-2013).”18 The flow time is measured using the funnel shown in Fig. 1. The flow time of water glass was controlled by adding pure water in an effort to attain the same flow time of mortar matrix as that of the target HPFRCC. The mixture proportion of the target HPFRCC is listed in Table 2. This proportion is selected for the tension test specimens, as explained in a later section. The measured flow time of the mortar matrix was 36 seconds on average for each of the eight mixture batches with the same mixture proportion. The water glass to the pure water weight ratio was chosen to be 12:1 at a temperature of 25°C (77°F). The density of the water glass solution was 1.62 g/cm3 (101 lb/ft3), which is smaller than 1.89 g/cm3 (118 lb/ft3) of mortar matrix used for the tension test specimens. The color of PVA fibers, which is almost white, makes it difficult to distinguish them from the water glass solution. ACI Materials Journal/March-April 2016

Fig. 2—Mixing of fiber in water glass. Therefore, black-colored “target fibers” made from nylon were added to the matrix to simplify the image analysis. The mechanical properties of the target fibers are listed in Table 1. The volume fraction of the target fibers was set to 0.05% based on empirical trial-and-error results. The mixture states are shown in Fig. 2. The image analysis on target fibers (explained in a subsequent subsection) was conducted based on the assumption that these fibers flow in similar orientations as those associated with the PVA fibers. Simulation method Water glass solution containing PVA and the target fibers was poured into the mold, using the same method as that used for HPFRCC casting. The mold was constructed with transparent acrylic plates. For simulations of the flow in the tension test specimens, the cross section of the mold was chosen to be 40 x 40 mm (1.57 x 1.57 in.) to be over three times the fiber length of 12 mm (0.47 in.), considering 3-D orientation of fibers.10 The testing parameters included the casting direction and the volume fraction Vf of the PVA fibers. The dimensions of the mold are shown in Fig. 3. Two molds were prepared: one for the casting along the horizontal direction and a second for the casting along the vertical direction. Water glass solution was poured into the mold using a bucket at the points indicated by the arrows in Fig. 3. The pouring time was approximately 20 seconds and was similar in value to the case of casting of the tension test specimens. After pouring, photos of the x-y and z-x planes were taken using two digital cameras at in-plane resolution of 6000 x 4000 pixels. The setup of the cameras for the horizontal casting simulation is shown in Fig. 4. Simulated volume fractions of PVA fibers are 0.1%, 0.5%, 1.0%, 1.5%, and 2.0%. For each volume fraction, three image specimens were cast, followed by photo capturing. An example of the photograph (Vf = 0.1%, horizontal casting, z-x plane) is shown in Fig. 5. Image analysis and calculation of fiber angle Image analysis was conducted to obtain the fiber angles in the water glass solution. The image analysis and calculation of the fiber angles were carried out for the target fibers that occupied the central 40 mm (1.57 in.) region, as shown in Fig. 5. The procedure of the image analysis is described as follows: 1. The photograph is cropped to include only the target region (Fig. 6(a)); ACI Materials Journal/March-April 2016

Fig. 3—Mold for visualization simulation. (Note: 1 mm = 0.0394 in.)

Fig. 4—Camera setup (horizontal casting). 2. The image is binarized and the noise is filtered (Fig. 6(b)); 3. RGB values (red-green-blue values in bit for each color) of the pixel data are read with position coordinates (Xi, Yi); and 4. The sequences of black-colored pixels are grouped and labeled (Fig. 6(c)). After this process, a straight line approximation is calculated using the position coordinates of the pixels of the same group, using least-squares regression analysis by minimizing the distance between the point and the line. The fiber angle is defined as the angle between the fitted line and the longitudinal axis (x-axis). The fiber angle ranges between –90 and +90 degrees. Examples of fiber angle histograms (Vf = 0.1%, horizontal casting) are shown in Fig. 7. The diagram on the right side of this figure corresponds to the calculated histogram result based on the photograph of Fig. 5, and the analysis methodology shown in Fig. 6. As indicated in Fig. 6(b), fiber angles mostly range between 0 to 45 degrees. The frequencies of the fiber angles that are extracted based on the three-time pouring and photography are added together, and one diagram is drawn for each parameter of the simulation test. All the fiber angle histograms are shown in Fig. 8. As expected, there is a tendency that the fibers flow along the longitudinal direction in the case of horizontal casting, 133

Fig. 5—Example of photograph (Vf = 0.1%, horizontal, z-x). (Note: 1 mm = 0.0394 in.)

Fig. 6—Image analysis procedures. (Note: 1 mm = 0.0394 in.)

Fig. 7—Examples of fiber angle histograms. and along the perpendicular direction in the case of vertical casting. The presented solid lines and the respective values of the diagrams are explained in the next section. PROBABILITY DENSITY FUNCTION FOR FIBER ORIENTATION DISTRIBUTION Approximation based on elliptic function For the purpose of quantitative evaluation of the fiber orientation distribution, an approximation methodology using the elliptic function is introduced. This methodology was studied in the field of “Japanese traditional paper (Washi).”19 The relative frequency for each class of fiber angle is transformed into a vector with the argument set to be equal to the fiber angle, as shown in Fig. 9. The trajectory traced by the terminal points of these vectors is approximated by an ellipse fitted using the least-squares method. The ellipse is expressed as a function of two radii, a and b, and the angle with respect to the x*-axis, θr, as 134

shown in the figure. The value of θr ranges between –45 and +45 degrees, and the argument of radius a corresponds to θr. As shown in Fig. 9, a random fiber orientation results in a circle, whereas the orientation tendency of the fibers along the longitudinal direction results in an ellipse. As the longitudinal directionality becomes greater, the shape of the ellipse becomes narrower. The ratio of the two radii, k = a/b, can express the shape of the ellipse. This ratio of two radii is defined as “orientation intensity”, and the angle θr is defined as “principal orientation angle”. The orientation intensity value reflects the orientation tendency of the fibers that lie along the principal orientation angle. When the fibers orient perfectly randomly, k is equal to 1. As shown in Fig. 10, when the fibers show an increased directional orientation toward the principal orientation angle, the value of k is larger than 1. In contrast, when the fibers orient perpendicularly with respect to the principal orientation angle, the value of k is smaller than 1. ACI Materials Journal/March-April 2016

Fig. 8—All fiber angle histograms.

Fig. 10—Definition of principal orientation angle and orientation intensity.

Fig. 9—Approximation method based on elliptic function. The PDF that expresses the relative frequency corresponding to the fiber angle θ is described by Eq. (1): hereafter, the PDF is referred to as “elliptic distribution”. The parameters for this function are the orientation intensity k and the principal orientation angle θr. When θr is equal to zero, the elliptic function is simply given by Eq. (5). The definite integral calculus of Eq. (1) and Eq. (5) in –π/2 ≤ π/2 gives 1 (the sum of probability).

p (θ) =

k C ⋅ (1) 2 π cos θ + A sin θ cos θ + B sin 2 θ

ACI Materials Journal/March-April 2016



A=

(1 − k ) sin 2θ r (2) 1 + (k − 1) sin 2 θ r



B=

k − (k − 1) sin 2 θ r (3) 1 + (k − 1) sin 2 θ r



C=

1 (4) 1 + (k − 1) sin 2 θ r



p (θ) =

1 k ⋅ (5) π cos 2 θ + k ⋅ sin 2 θ

Approximation of visualization simulation results The results of the approximation of the fiber angle distribution obtained in the visualization simulation are shown 135

in Fig. 8 by solid lines. The values of the orientation intensity k and the principal orientation angle θr are also listed in the figures. When the directionality of the fiber orientation increases along the longitudinal direction, the value of the orientation intensity is over 5 (cases of z-x plane for Vf = 1.0%, 1.5%, and 2.0%). In vertical casting, the fiber angles tend to align along the perpendicular direction, and there are the cases that the value of the orientation intensity becomes smaller than 0.5 (cases of x-y and z-x planes for Vf = 1.5% and 2.0%). These evaluations are done for two planes individually. The estimated probabilities for each plane are multiplied to express the probability in 3-D orientation in a later section. UNIAXIAL TENSION TEST Test outline For verification of the influence of fiber orientation on tensile behavior, the uniaxial tension test was conducted. As explained in the Introduction, PVA fibers with a diameter of 0.10 mm (3.9 × 10–3 in.) are used to observe the bridging law directly, subjected to single crack formation. The mechanical properties of the PVA fiber are listed in Table 1. The fibers used for the tension test are same as those used in the visualization simulation. The mixture proportion of mortar matrix has already been presented in Table 2. The fiber volume fraction is 2.0%. The testing parameter is the casting direction along both the horizontal and vertical directions. Two types of molds for each casting direction were prepared, as shown in Fig. 11. The matrix with fibers was poured into the mold using a bucket employing the same approach as the one used for the visualization simulation. The pouring time was controlled to be approximately 20 seconds in the test region. The dimensions of the specimen and the specimen setup are shown in Fig. 12. The cross section of the test region is 50 x 50 mm (1.97 x 1.97 in.) square to be over three times the fiber length of 12 mm (0.47 in.), considering 3-D orientation of fibers.10 The total length of the specimen is 510 mm (20.1 in.). A 2000 kN (450 kip) universal loading machine was used. Pin-fix ends were used at the boundaries to minimize possible effects of development of external moment because of setup irregularity, and secondary moment influencing local fracture.9 The carbon fiber sheets were attached at both ends to avoid peel-off of the steel plate. Measurement items were tensile load and deformation in the test region using two pi-type displacement transducers. Two series of test in different period (Batch No. 1, compressive strength = 39.2 N/mm2 [5.69 ksi]; and Batch No. 2, compressive strength = 41.0 N/mm2 [5.95 ksi]) were carried out. Test results All specimens fractured by a single crack. Some of the specimens had a fine crack before loading because of an unskillful treatment during the formwork removal. The test results of these specimens are excluded from the following discussions. The curves of the tensile stress-crack width are shown in Fig. 13. It is clearly recognized that the casting direction remarkably affects the tensile performance. The test results are summarized in Table 3. The average tensile 136

Fig. 11—Molds for tensile test specimen.

Fig. 12—Tensile test specimen. (Note: 1 mm = 0.0394 in.) stress at the maximum load after the sudden drop of the load (second peak) is 3.51 and 1.67 N/mm2 (0.509 and 0.242 ksi) for the horizontal casting and the vertical casting specimens, respectively. The tensile stress of the second peak of horizontal casting specimens is more than two times that of the vertical casting specimens. The crack width at the second peak of the horizontal casting specimens is, on average, 1.73 times higher than the corresponding value of the vertical casting specimens. Characteristic example photographs of the fractured surface after loading are shown in Fig. 14. It is clearly seen that the protruded fibers from the surface of the horizontal casting specimen are many more and longer than those of the vertical casting specimen. BRIDGING LAW CONSIDERING FIBER ORIENTATION Trilinear model for pullout load versus crack width relationship The calculations of the bridging law of the PVA fiber considering the fiber orientation distribution are conducted. The elliptic distribution expressed by the orientation intensity and the principal orientation angle is adopted for the PDF estimation of the fiber orientation distribution. The orientation intensity and the principal orientation angle used

ACI Materials Journal/March-April 2016

Fig. 13—Tensile stress-crack width curve. Table 3—Tension test results At cracking (first peak) Casting direction

Horizontal

Fig. 14—Fracture surface after loading. (Note: 1 mm = 0.0394 in.) for calculations are based on these results of the visualization simulations. The pullout properties of the single fiber are required to calculate the bridging law. Several researchers have studied the bond behavior of PVA single fiber to cementitious matrix.20-23 Table 4 lists the results from previously published studies in which the pullout tests of the single fiber were performed. It has been known that the bond behavior of the PVA fiber consists of two stages—that is, the chemical bond stage and the friction stage.20 The pullout load-displacement relationships of the PVA fiber commonly exhibit the first peak in the debonding process of the chemical bond, and slip hardening or softening.21 Table 4 lists the information of the matrix used, the fiber diameter, first peak load, and the maximum load in the friction process (second peak). Based on these results, a trilinear model is assumed to express the relationship between the pullout load and the crack width for a single fiber, as shown in Fig. 15. The pullout load for the first branch, Pa, and for the maximum, Pmax, corresponds to the first peak load and the second peak load, respectively. As seen in Table 4, there is no test result listed on PVA fibers with a diameter of 0.10 mm (3.9 × 10–3 in.), as used in this study. Furthermore, the water-cement ratio (w/c) used in this study is 0.56, which also differs from corresponding ratio values in prior studies. Considering the differences of fiber diameters and mixture proportions of the matrix, the values of Pa and Pmax are assumed to be 1.5 and 3.0 N (0.34 and 0.67 lbf), respectively. These values correspond to the values of 0.24 and 0.48 N (0.054 and 0.108 lbf) for the same tensile stress of a PVA fiber with a diameter of 0.04 mm (1.6 × 10–3 in.). The values of 0.24 and 0.48 N (0.054 and 0.108 lbf) are in the ranges of the test results reported by Kiyota et al.22 ACI Materials Journal/March-April 2016

Maximum after cracking (second peak)

Tensile stress, N/ mm2 (ksi)

Crack width, mm (in.)

Tensile stress, N/ mm2 (ksi)

Crack width, mm (in.)

TH20-1*

4.49 (0.651)

0.032 (0.0013)

3.70 (0.537)

0.460 (0.0181)

TH20-2*

4.41 (0.640)

0.034 (0.0013)

3.85 (0.558)

0.463 (0.0182)

TH20-3†

3.17 (0.460)

0.030 (0.0012)

2.97 (0.431)

0.446 (0.0176)

Average

4.02 (0.583)

0.032 (0.0013)

3.51 (0.509)

0.456 (0.0180)

TV20-1*

3.53 (0.512)

0.023 (0.0009)

1.37 (0.199)

0.328 (0.0129)

TV20-2†

2.35 (0.341)

0.013 (0.0005)

1.55 (0.225)

0.177 (0.0070)

TV20-3†

3.53 (0.512)

0.030 (0.0012)

2.09 (0.303)

0.284 (0.0112)

Average

3.14 (0.455)

0.022 (0.0009)

1.67 (0.242)

0.263 (0.0104)

ID

Vertical

Batch No. 1, compressive strength = 39.2 N/mm2 (5.69 ksi).

*

Batch No. 2, compressive strength = 41.0 N/mm2 (5.95 ksi).

†

The crack widths, δa and δmax, are those corresponding to the loads of Pa and Pmax, respectively. These crack widths correspond to the slip-out displacements at the first and second peak loads in the pullout test. The slip-out displacements for the two peaks are simply assumed to be 0.1 and 0.3 mm (3.9 × 10–3 and 12 × 10–3 in.) from the test results of Yang et al.23 The crack width becomes twice the slip-out displacement before the maximum load because the fiber slips out from the both embedded sides. When the pullout load starts to decrease at the short embedded side of the fiber, the slip-out displacement at the long embedded side decreases.24 To express this phenomenon using a simple trilinear model, the crack width at the maximum load is assumed to be 1.5 times the slip-out displacement at the second peak in the pullout test. As a result, the values of δa and δmax are assumed to be 0.2 and 0.45 mm (7.8 × 10–3 and 18 × 10–3 in.), respectively. The softening branch is decided as the pullout load becomes zero, when the crack width equals the embedded length of the short side of the fiber, lb. These assumed values are summarized in Table 5 and illustrated in Fig. 15. 137

Table 4—Previous PVA fiber pullout test results Researcher

w/c

Fiber diameter, mm (in.)

0.27 Kanda et al.

20

0.42

0.014 (0.55 × 10 ) –3

0.62 Redon et al.21

0.30

0.044 (1.7 × 10–3)

0.34 Kiyota et al.

22

0.42

0.038 (1.5 × 10 ) –3

0.62 Yang et al.23

0.58

0.039 (1.5 × 10–3)

First peak load, N (lbf)

Second peak load, N (lbf)

0.05 to 0.25 (0.011 to 0.056)

—

0.12 to 0.20 (0.027 to 0.045)

—

0.07 to 0.14 (0.016 to 0.031)

—

0.8 to 1.2 (0.18 to 0.27)

1.1 to 1.6 (0.25 to 0.36)

0.3 to 0.6 (0.07 to 0.13)

0.5 to 1.3 (0.11 to 0.29)

0.4 to 0.6 (0.09 to 0.13)

0.4 to 1.3 (0.09 to 0.29)

0.2 to 0.4 (0.04 to 0.09)

0.4 to 0.9 (0.09 to 0.20)

0.3 to 0.6 (0.07 to 0.13)

0.5 to 1.0 (0.11 to 0.22)

P, is given by Eq. (7), expressing the snubbing effect and the fiber strength degradation. σbridge (δ ) =



=

Pbridge (δ )



Am Vf Af

⋅ ∑ ∑ ∑ Pij (δ, ψ ) ⋅ pxy (θi ) ⋅ pzx (φ j ) ⋅ px ( yh , zh ) ⋅ ∆θ ⋅ ∆φ ⋅ ( ∆y ⋅ ∆z ) h

j

i

(6)

Fig. 15—Trilinear model for pullout load. (Note: 1 mm = 0.0394 in.) Bridging law simulation method The bridging stress can be obtained as the total pullout load of fibers divided by the cross-sectional area of the matrix. Moreover, the elliptic distribution is adopted to express the fiber orientation distribution. The snubbing effect24 and the fiber strength degradation20 are also considered in this study. The snubbing effect exhibits the increment of the pullout load of the fiber due to the edge reaction, when the fiber has the angle with the normal direction of cracking plane. The fiber strength degradation has been adopted for the polymeric fibers, which strength decreases when the fiber is pulled out slantingly from its embedded direction. The definitions of the coordinate system and the fiber angle in consideration of the snubbing effect and the fiber strength degradation are shown in Fig. 16. The fiber angles, θ and ϕ, are the angles between the x-axis and the projected lines of the fiber (angle of ψ to x-axis) to x-y and z-x planes, respectively. When the angle ψ increases, the pullout load also increases, owing to the snubbing effect. However, as this angle increases, fiber strength decreases, and the fiber ruptures easily (Fig. 15). The elliptic distribution is considered for each of the x-y and z-x planes. Therefore, the formula expressing the bridging stress can be given by Eq. (6). Equation (6) is derived by the summation of the pullout load of the fibers that exist in bridging the crack surface with the probability given in the elliptic distribution. The probabilities for x-y and z-x planes are multiplied to express the probability in 3-D orientation. The pullout load of a single fiber,

138

P = Ppull · e f·ψ < Prup · e–f ′·ψ (once exceeded, P = 0) (7)

where σbridge is bridging stress; δ is crack width; Pbridge is bridging force (= total of pullout load); Am is cross-sectional area of the matrix; Vf is fiber volume fraction; Af is cross-sectional area of a fiber; P is pullout load of a single fiber; Ppull is pullout load of a single fiber at a zero fiber angle; Prup is pullout load of a single fiber at rupture at a zero fiber angle; f is snubbing coefficient; f ′ is fiber strength reduction factor; pxy, pzy are probability, based on elliptic distribution; px is probability of fiber distribution along x-axis; ψ is fiber angle to x-axis; θ is angle between x-axis and projected line of the fiber to x-y plane; and ϕ is angle between x-axis and projected line of the fiber to z-x plane The PDF, px (y, z), gives the probability for the existence of the fiber in the x-axis direction. In this study, px (y, z) is assumed to be constant. This means that the fibers are randomly distributed along the longitudinal direction of the specimen. The input values for the parameters are listed in Table 5. The orientation intensities for the horizontal casting are selected to be 1.5 and 6 for the x-y and the z-x planes, respectively. On the other hand, the corresponding values for the vertical casting are set to 0.5. These values are chosen based on the results of the visualization simulation for Vf = 1.5% and 2.0% (Fig. 8). The principal orientation angles are set to zero for calculation simplification. This value almost agrees with the average value of all the results of the visualization simulation. The calculations were done by using spreadsheet software. Comparison with tension test result The calculated curves showing the variation of the bridging stress (tensile stress) with the crack width are shown in Fig. 17 together with the tension test results, for both ACI Materials Journal/March-April 2016

Table 5—Parameters for bridging law Parameter

Input value

Remarks

First peak load Pa, N (lbf)

1.5 (0.34)

*

Crack width at Pa, δa, mm (in.)

0.2 (7.8 × 10–3)

0.1 mm (3.9 × 10–3 in.)* × 2

Maximum load Pmax, N (lbf)

3.0 (0.67)

*

Crack width at Pmax, δmax, mm (in.)

0.45 (18 × 10–3)

0.3 mm (12 × 10–3 in.)* × 1.5

Fiber strength σfu, N/mm2 (ksi)

774 (112)

1200 N/mm2 (174 ksi) × 0.645†

Snubbing coefficient f

0.5

†

Fiber strength reduction factor f ′

0.3

†

x-y plane

Orientation intensity kxy

z-x plane

Orientation intensity kzx

Horizontal casting

1.5

Vertical casting

0.5

Value near to Vf 1.5% and 2.0% visualizations

0

For calculation simplification‡

Horizontal casting

6

Vertical casting

0.5

Value near to Vf 1.5% and 2.0% visualizations

0

For calculation simplification‡

Principal orientation angle θr,xy

Principal orientation angle θr,zx

*

Assumed value based on Kiyota et al. and Yang et al.

†

Assumed value for PVA fiber by Kanda et al.5

‡

Approximately average value of all Vf visualizations.

22

23

Notes: PVA fiber is: 0.10 mm (3.9 × 10–3 in.) in diameter, 12 mm (0.47 in.) in length.

Fig. 16—Definitions of coordinate system and fiber angle. the horizontal and vertical casting specimens. Because the calculated curves exhibit the bridging stress by fibers after cracking, the elastic region before cracking in the tension test (indicated by dotted line) cannot be compared with the calculated curve. The calculated curves express well the test results after the first peak in the tension test. Based on these calculations, the only parameter that differs between the horizontal and vertical casting is the orientation intensity. The difference of the fiber orientation intensity identifies a clear influence on the bridging law. Fiber effectiveness is also defined to express the effectiveness of the fiber in bridging the crack surface. It is calculated as the ratio of the number of fibers crossing the crack surface (neither slipping out nor rupturing) to the theoretical number of total fibers in a unit volume. The fiber effectiveness is equal to the orientation factor at a crack width of zero. Figure 17 also shows the calculation results of fiber effectiveness and crack width relationship both for the horizontal and the vertical casting specimens. The fiber effectiveness values at a crack width of zero are 0.544 and 0.315 for horizontal and vertical casting, respectively. The difference of the fiber orientation distribution causes this disparity. The fiber effectiveness decreases as the crack width increases because of slipping out or because of fiber rupture. The “step” can be seen on the curve, when the fibers ACI Materials Journal/March-April 2016

rupture more frequently. The balance between the increase of the pullout load and the loss of the bridging because of fiber rupture leads to the maximum bridging stress. After the end of the “step”, the fiber effectiveness of horizontal and vertical casting becomes 0.346 and 0.116, respectively. These values are considered almost equal to the orientation factor after fracture, that is, the ratio of the fibers that slipped out from the crack surface. The tension test results shown in Fig. 14 support this consideration. CONCLUSIONS To investigate the influence of the fiber orientation distribution on the bridging performance in PVA-FRCC, visualization simulation using water glass solution and calculation of the bridging law considering the fiber orientation distribution were conducted. The main parameter of the investigations is the casting direction of FRCC. The followings are concluded from this study. 1. From the visualization simulation, the fibers have a tendency to flow along the longitudinal direction in the case of horizontal casting and along the perpendicular direction in the case of vertical casting. 2. To evaluate the fiber orientation distribution quantitatively, a new approximation methodology using an elliptic function was introduced. The PDF named elliptic distribution is characterized by the principal orientation angle and the orientation intensity. 3. From the visualization simulation, while the value of the orientation intensity shows over 5 in the case of horizontal casting, there are the cases that the orientation intensity becomes smaller than 0.5 in the case of vertical casting. 4. The bridging stress versus crack width relationship was calculated considering the elliptic distribution, the snubbing effect, and the fiber strength degradation. The calculated bridging curves were compared with the results of the 139

Fig. 17—Calculated bridging law and fiber effectiveness. tension test in which the specimens were fabricated by horizontal and vertical casting. The calculated curves expressed the test results after first cracking well. 5. The differences of the fiber orientation distribution clearly indicated an influence on the bridging law. Based on the calculation results for the bridging law, it was considered that the balance between the increasing pullout load and the loss of the bridging force because of the fiber rupture leads to the maximum bridging stress. FUTURE RESEARCH In this study, only one type of mold and matrix was used for the visualization simulation. It is considered that the other factors such as casting method, fresh-state properties, flow, vibration, and formwork geometry also have influences to the fiber orientation. It is necessary that the influence of these factors on the principal orientation angle and orientation intensity be clarified. In addition, the influence of the fiber diameter variation to the fiber orientation distribution should be investigated. Further experiments of the flow simulation are necessary to study the adaptability of the proposed PDF. If the fiber orientation can be evaluated more quantitatively, the tensile characteristics of FRCC can be estimated more precisely. AUTHOR BIOS

ACI member Toshiyuki Kanakubo is an Associate Professor at the Department of Engineering Mechanics and Energy, University of Tsukuba, Tsukuba, Japan, where he received his PhD. His research interests include high-performance fiber-reinforced cementitious composites (HPFRCCs), structural behavior of fiber-reinforced polymer reinforced concrete structures, and bond properties of reinforcement and concrete. Masaru Miyaguchi is a Student in the master’s program at the Department of Engineering Mechanics and Energy, University of Tsukuba, where he received his BE. His research interests include the evaluation of fiber orientation of, and its influence on, high-performance fiber-reinforced cementitious composites (HPFRCCs).

140

Kohei Asano is a Research Associate at the Department of Architecture, Miyakonojo College, National Institute of Technology, Miyakonojo, Japan. He received his PhD from the University of Tsukuba. His research interests include the mechanical behavior of high-performance fiber-reinforced cementitious composites (HPFRCCs), and structural performance of reinforced concrete members using HPFRCC.

ACKNOWLEDGMENTS

The authors wish to express their gratitude and sincere appreciation to the Kuraray Co., Ltd., for providing the PVA fiber. The tension test was performed in cooperation with R. Tsukizaki, a former student in the master’s program of the University of Tsukuba. This study was supported by the JSPS KAKENHI Grant Number 24656319.

REFERENCES

1. Bentur, A., and Mindess, S., Fiber Reinforced Cementitious Composites, second edition, Taylor & Francis, London, UK, 2007, 601 pp. 2. Balaguru, P., and Shah, S. P., Fiber Reinforced Cement Composites, McGraw Hill, New York, 1992, 530 pp. 3. Naaman, A. E., and Reinhardt, H. W., “Characterization of High Performance Fiber Reinforced Cement Composites—HPFRCC,” High Performance Fiber Reinforced Cement Composites 2 (HPFRCC2), RILEM Proceedings No. 31, E&FN Spon, London, UK, 1995, pp. 1-24. 4. Li, V. C., “From Micromechanics to Structural Engineering—The Design of Cementitious Composites for Civil Engineering Applications,” Structural Engineering/Earthquake Engineering, V. 10, No. 2, 1993, pp. 37-48. 5. Kanda, T., and Li, V. C., “Effect of Fiber Strength and Fiber-Matrix Interface on Crack Bridging in Cement Composites,” Journal of Engineering Mechanics, ASCE, V. 125, No. 3, 1999, pp. 290-299. doi: 10.1061/ (ASCE)0733-9399(1999)125:3(290) 6. Li, V. C., and Leung, C. K. Y., “Steady-State and Multiple Cracking of Short Random Fiber Composites,” Journal of Engineering Mechanics, ASCE, V. 118, No. 11, 1992, pp. 2246-2264. doi: 10.1061/ (ASCE)0733-9399(1992)118:11(2246) 7. Rokugo, K., and Kanda, T., eds., “Strain Hardening Cement Composites: Structural Design and Performance,” State-of-the-Art Report of the RILEM Technical Committee 208-HFC, SC3, Springer, 2013, 90 pp. 8. Laranjeira, F.; Aguado, A.; Molins, C.; Grünewald, S.; Walraven, J.; and Cavalaro, S., “Framework to Predict the Orientation of Fibers in FRC: A Novel Philosophy,” Cement and Concrete Research, V. 42, No. 6, 2012, pp. 752-768. doi: 10.1016/j.cemconres.2012.02.013 9. Kanakubo, T., “Tensile Characteristics Evaluation Method for Ductile Fiber-Reinforced Cementitious Composites,” Journal of Advanced Concrete Technology, V. 4, No. 1, 2006, pp. 3-17. doi: 10.3151/jact.4.3

ACI Materials Journal/March-April 2016

10. Li, V. C., and Wang, S., “On High Performance Fiber Reinforced Cementitious Composites,” JCI Proceedings of the Symposium on Ductile Fiber-Reinforced Cementitious Composites, 2003, pp. 13-23. 11. Naaman, A., “A Statistical Theory of Strength for Fiber Reinforced Concrete,” PhD dissertation, Massachusetts Institute of Technology, Cambridge, MA, 1972, 196 pp. 12. Stroeven, P., “Stereological Principles of Spatial Modeling Applied to Steel Fiber-Reinforced Concrete in Tension,” ACI Materials Journal, V. 106, No. 3, May-June 2009, pp. 213-222. 13. Dupont, D., and Vandewalle, L., “Distribution of Steel Fibres in Rectangular Sections,” Cement and Concrete Composites, V. 27, No. 3, 2005, pp. 391-398. doi: 10.1016/j.cemconcomp.2004.03.005 14. Xia, J., and Mackie, K., “Axisymmetric Fiber Orientation Distribution of Short Straight Fiber in Fiber-Reinforced Concrete,” ACI Materials Journal, V. 111, No. 2, Mar.-Apr. 2014, pp. 133-141. doi: 10.14359/51686721 15. Liu, J.; Li, C.; Liu, J.; Cui, G.; and Yang, Z., “Study on 3D Spatial Distribution of Steel Fibers in Fiber Reinforced Cementitious Composites through Micro-CT Technique,” Construction and Building Materials, V. 48, 2013, pp. 656-661. doi: 10.1016/j.conbuildmat.2013.07.052 16. Torigoe, S.; Saito, T.; Horikoshi, T.; Hamada, T.; and Ogawa, A., “Study on Evaluation Method for PVA Fiber Distribution in Engineered Cementitious Composite,” Proceedings of the JCI International Workshop on Ductile Fiber Reinforced Cementitious Composites, 2001, pp. 95-101. 17. Kanda, T.; Tomoe, S.; Nagai, S.; Maruta, M.; Kanakubo, T.; and Shimizu, K., “Full Scale Processing Investigation for ECC Pre-Cast Structural Element,” Journal of Asian Architecture and Building Engineering, V. 5, No. 2, 2006, pp. 333-340. doi: 10.3130/jaabe.5.333

ACI Materials Journal/March-April 2016

18. Japan Society of Civil Engineers, “Standard Specifications for Concrete Structures — 2013, Test Methods and Specifications,” JSCE Standard, 2013, 2013, pp. 281-282. (in Japanese) 19. Enomae, T.; Han, Y. H.; and Isogai, A., “Fiber Orientation Distribution of Paper Surface Calculated by Image Analysis,” Proceedings of International Papermaking and Environment Conference, Book 2, Tianjin, P.R. China, 2004, pp. 355-368. 20. Kanda, T., and Li, V. C., “Interface Property and Apparent Strength of High-Strength Hydrophilic Fiber in Cement Matrix,” Journal of Materials in Civil Engineering, ASCE, V. 10, No. 1, 1998, pp. 5-13. doi: 10.1061/ (ASCE)0899-1561(1998)10:1(5) 21. Redon, C.; Li, V. C.; Wu, C.; Hoshiro, H.; Saito, T.; and Ogawa, A., “Measuring and Modifying Interface Properties of PVA Fibers in ECC Matrix,” Journal of Materials in Civil Engineering, ASCE, V. 13, No. 6, 2001, pp. 399-406. doi: 10.1061/(ASCE)0899-1561(2001)13:6(399) 22. Kiyota, M.; Mihashi, H.; Kanda, T.; and Kawamata, A., “Study on Bond Characteristics of Fibers in Cementitious Composites,” JCI Proceedings of the Japan Concrete Institute, V. 23, No. 2, 2001, pp. 187-192. (in Japanese) 23. Yang, E. H.; Wang, S.; Yang, Y.; and Li, V. C., “Fiber-Bridging Constitutive Law of Engineered Cementitious Composites,” Journal of Advanced Concrete Technology, V. 6, No. 1, 2008, pp. 181-193. doi: 10.3151/jact.6.181 24. Li, V. C.; Wang, Y.; and Backer, S., “A Micromechanical Model of Tension-Softening and Bridging Toughening of Short Random Fiber Reinforced Brittle Matrix Composites,” Journal of the Mechanics and Physics of Solids, V. 39, No. 5, 1991, pp. 607-625. doi: 10.1016/0022-5096(91)90043-N

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ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M14

Strain Rate Sensitivity of Fiber-Reinforced Cementitious Composites by H. Othman and H. Marzouk An experimental investigation has been conducted to determine the effects of strain rate on fiber-reinforced cementitious composite (FRCC) matrixes. Compressive strength, modulus of elasticity, and flexural tensile strength are investigated under various strain rates ranging from the static to the seismic and/or impact level. Three different matrixes with compressive strengths ranging from 80 to 130 MPa (12 to 19 ksi) are investigated. The first matrix is without fiber, while the other two contain 2% straight steel fibers by volume. The tests are carried out according to ASTM standards. The dynamic increase factor (DIF) formulation recommended by the European CEB-fib is described. Experimental results showed that the rate sensitivity decreases with an increase in the matrix compressive strength. Additionally, it has been found that CEB-fib Model Code 2010 fits well with high-strength concrete. On the other hand, the CEB-fib Model (2010) overestimates both compressive and tensile strengths enhancement for FRCC with compressive strength over 110 MPa (16 ksi). Keywords: dynamic increase factor; fiber-reinforced cementitious composites (FRCC); impact loading; quasi-static; steel fiber; strain rate effect.

INTRODUCTION Recently, there has been a growing realization that important structures should be designed to resist both static and dynamic loads. The material required to construct such types of structures should have enhanced static and dynamic properties. Fiber-reinforced cementitious composite (FRCC) materials seem to be the best choice to fit needed properties for many structures. Such structures include: transportation structures, offshore structures, protective structures, and aircraft launching platforms. FRCC has enhanced dynamic properties, especially under impact resistance loads. It has a high resistance to spalling, scabbing, and fragmentation, and high energy absorption capacity. The use of FRCC in impact/blast-resistant structures, especially ultra-highperformance fiber-reinforced concrete (UHP-FRC) is increasing. However, there are insufficient studies to fully describe the dynamic behavior of FRCC.1 Therefore, there is an urgent need to develop a better understanding of the dynamic response, and the nonlinear behavior of FRCC members subjected to dynamic loading. For materials subjected to dynamic effects such as impact loading response over a relatively short time period, the strain rates reach magnitudes considerably higher than that of static conditions. Figure 1 shows typical orders of magnitude of strain rates for different loading types. It is well known that high strain rates result in increased mechanical properties in most materials.2 Although the reason for this enhancement is not entirely understood, it is widely considered to be a material property. Dynamic increase factor (DIF) is the most popular method for taking ACI Materials Journal/March-April 2016

account of strain rate effects on both deformation and failure.2 The DIF is defined as the ratio of the dynamic to static strength. DIFs is of direct use in finite element modeling and analysis of reinforced concrete structures subjected to dynamic loading conditions.2,3 Abrams was the first researcher who, in 1917, observed the effect of changing strain rates on concrete response.2,4 Further, numerous experimental studies have demonstrated that the rate effect on strength, modulus of elasticity, strain, and fracture energy of concrete.3-7 Most of strain rate studies have been conducted on plain normal and high-strength concretes. Such studies typically proposed models to be used to estimate the concrete DIFs at certain strain rate. These models are mainly functions in concrete compressive strength, and quasi-static and dynamic strain rate.3 Although there are some differences in estimated values at certain strain rates using these models, all these studies typically concluded that: 1) the stiffness and strength properties of concrete increase significantly under high strain rates; 2) DIFs are higher for concretes with lower strengths; and 3) the strength enhancement is different for compression and tension.3,4 On the other hand, the increase in the modulus of elasticity and the peak strain corresponding to the peak stress is relatively small.3 The effect of fibers is similar over various fibers types. Fibers have little effect on compressive properties.8 On the other hand, fibers enhance significantly tensile/flexural, shear, and ductility properties.8,9 Fiber-reinforced concrete (FRC) exhibits enhanced impact resistance compared to plain concrete.4,10 Different conclusions have been drawn for strain rate effect on FRC materials. Gopalartnam and Shah11 and Maalej et al.12 concluded that FRC is more rate-sensitive than plain concrete. On the other hand, Millard et al.’s8 results showed that the DIF is greater for specimens without fibers and decrease with the increase of fiber contents. The rate sensitivity of fiber-matrix interface or pullout of short straight fibers has been shown to be independent of strain rate.13,14 In the present study, the commonly used cementitious materials—HSC and FRCC—are investigated. Additionally, the study is mainly focused on strain rate range from static to seismic or low-velocity impact (Fig. 1), because this strain domain most relevant to common load cases on civil engineering structures. ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2014-395.R3, doi: 10.14359/51688461, received July 18, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

143

RESEARCH SIGNIFICANCE The objective of the current investigation is to develop a fundamental understanding of strain rate effect on the behavior of FRCC material. The influence of matrix strength is studied. Aspects investigated included compressive strength, modulus of elasticity, and flexural tensile strength at six different strain rates ranging from the static to the seismic and/or impact level. This investigation is a part of an ongoing research program by the authors focusing in experimental study as well as numerical modeling of HSC and FRCC slabs subjected to impact. This investigation was motivated with the lack of DIF models that can be used in finite element numerical simulation of impact load conditions for FRCC materials. DIFs obtained from this paper will be implemented in a material constitutive model for concrete in an explicit finite element code. DIF OF CEB-fib CONCRETE MODEL (2010) The most comprehensive formulas for predicting the strain rate enhancement of concrete are presented by the CEB-fib Model Code (Comite Euro-International du Beton-Federation Internationale de la Precontrainte). The CEB-fib Model Code (2010) formulas are based on the 1988 CEB Bulletin 187.3 The CEB Bulletin 187 itself is based on work by Reinhardt in 1985.3 The provisions of the CEB-fib Model Code (2010)15 covers concretes up to a characteristic strength of 120 MPa (18 ksi), including new fiber-reinforced cementitious materials. CEB-fib proposes a series of strain-ratedependent relationships for concrete in both compression and tension. These relationships are independent of concrete material properties and are applicable for strain rate up to 3 × 102 s–1. DIFs formulas of the CEB-fib Model Code 2010 are summarized in Fig. 2. In general, DIF is related to a basic static strength measured at a specific quasi-static strain rate. In the literature, this

Fig. 1—Typical strain rates for various types of loading and present investigation domain.13

strain rate is varied from 1 × 10–8 to 1 × 10–5 s–1.3 Because these experiments would be fitted with CEB-fib Model Code 2010 formulas, the quasi-static strain rates recommended by CEB-fib are chosen; quasi-static strain rates of 3 × 10–5 and 1 × 10–6 s–1 are adapted for the reported experimental compressive and tensile/flexure tests, respectively.15 EXPERIMENTAL INVESTIGATION An experimental program is conducted to investigate compressive strength, elastic modulus, and flexural strength of HSC and FRCCs at six different strain rates ranging from static to the seismic and/or impact level. Three different mixture designs with target compressive strengths ranging from 80 to 120 MPa (12 to 17 ksi) are investigated. Identical specimens are used in both static and dynamic tests with similar loading and support conditions to avoid size effect. All tests are carried out at age of 56 days to allow materials reach their maximum strength. Two types of specimens (cylinders and prisms) are used for compressive and flexural strengths tests, respectively, leading to six series of tests. Materials Three different matrixes have been investigated. The first is a conventional, nonfibrous, high-strength concrete (HSC) with a target 56-day compressive strength of 80 MPa (12 ksi). This matrix includes 6% silica fume and is based on the composition developed by Marzouk.16 The second and third mixtures are FRCC containing 2% steel fiber by volume with target 28-day compressive strengths of 100 and 120 MPa (14.5 and 17 ksi), respectively. These matrixes are resulted from a series of trial mixtures and modifications on the composition developed by Rossi et al.17 Mixture modifications have been conducted at Ryerson University materials laboratory to reach the locally available materials in Canada. Table 1 provides the mixture composition for the three matrixes. Commonly used straight, smooth, high-strength steel fibers are used in the FRCC mixtures. These fibers have a small diameter of 0.2 mm (0.008 in.) and are 13 mm (0.5 in.) long. This fiber geometry has an aspect ratio of 65, offering a trade-off between good workability and high pullout resistance.9 Fiber content of 2% by volume is used in this investigation because this fiber content has been founded to be the optimum out of thousands of tests with high bending and direct tensile strength.18 Additionally, this fiber content is a commonly used percent in the industry. The fiber manu-

Fig. 2—Summary of concrete DIFs according to CEB-fib (2010). 144

ACI Materials Journal/March-April 2016

Fig. 3—Preparation and casting of cylinders with embedded fiber-optics sensors. Table 1—Mixture proportions by weight and target compressive strength Matrix Compressive strength* fc′, MPa (ksi) Portland cement Silica fume Fine sand (size < 0.5 mm) Constituent, kg/m (lb/yd ) 3

3

Coarse aggregate (size < 12 mm) Water High-range water-reducing admixture

HSC

FRCC1

FRCC2

80 (12)

100 (14.5)

120 (17)

450 (758)

960 (1618)

1050 (1770)

30 (51)

190 (320)

250 (421)

550 (927)

650 (1096)

630 (1062)

1100 (1852)

NA

NA

220 (371)

220 (371)

200 (337)

20 (34)

30 (51)

40 (67)

NA

156 (263)

156 (263)

Steel fibers 2% by volume *

Target compressive strength.

Notes: 1 mm = 0.0394 in.; NA is not available.

Compressive strength test procedures Compressive strength and elastic modules tests have been conducted on 100 x 200 mm (3.9 x 7.9 in.) cylinders. A hydraulic servo-controlled testing machine (MTS 815) is used to conduct the compression testing for both the quasi-static and dynamic ranges. For each tested matrix, three specimens are tested at each strain rate. Compressive tests are conducted according to ASTM C39 and the capture of the strain is completed according to ASTM C469. As shown in Fig. 3, the cylinders are equipped with an embedded fiber-optics sensor capable of measuring longitudinal deformations over a gauge length of 150 mm (6 in.). The embedded fiber-optics sensors are used to verify the displacement rate reading of the machine. More details

ACI Materials Journal/March-April 2016

Table 2—Summary of compressive strength and elastic modulus tests Matrix

Loading range

Strain rate, s–1

Loading rate, mm/min (in./min)

Quasi-static

3 × 10–5

0.36 (0.014)

3 × 10

3.60 (0.140)

3 × 10–3

36.0 (1.40)

1 × 10

120 (4.72)

3 × 10

360 (14.20)

1 × 10

1200 (47.20)

FRCC2

FRCC1

–4

HSC

facturer’s specified minimum tensile strength and elastic modulus of the fibers are 1900 MPa (275 ksi) and 205 GPa (29,730 ksi), respectively. All matrixes are mixed in a vertical axis shear mixer. Each mixture is cast in one batch with a size of 120 L (4.25 ft3) to have identical material properties for each matrix. No heat is used during casting or curing. All specimens are cured following the same procedures: under moist burlap and plastic for 1 week. Then, all specimens are taken out of their molds and stored in a moist-curing chamber at a temperature of 20oC (68°F) for an additional 3 weeks, then removed and placed to dry in laboratory air conditions until testing at the age of 56 days.

Dynamic range

–2 –2 –1

about fiber-optics sensors and the calibration process are given in Reference 19. The loading rate is set through the software on the controlling computer as the displacement rate. As listed in Table 2, there are three different series of compressive strength tests at six different strain rates. The adapted basic displacement rate for the first static test is 0.36 mm/min (0.0014 in./min) that corresponds to the quasistatic strain rate of 3 × 10–5 s–1. The highest loading rate used in this investigation is 1200 mm/min (47.20 in./min), which corresponds to strain rate of 10–1 s–1. This high strain rate can represent values of demand during seismic loading, or from vehicle impact on bridge piers.20 It is clear that the ratio of the highest to the lowest strain rate is 3300. This rate is sufficient for impact analysis; however, for strain rate corresponding to blast and explosion, a special Hoskins bar test must be used. 145

Flexural strength test procedures Three-point bending tests have been conducted on 100 x 100 x 400 mm (3.9 x 3.9 x 15.8 in.) prisms with a clear span of 300 mm (11.8 in.). Specimens are rotated 90 degrees from their casting position to reduce the effects of casting direction on the results. For each tested matrix, three specimens are tested at each strain rate. Testing and analysis of results have been carried out according to ASTM C1609. A second hydraulic servo-controlled (MTS 793) testing machine, shown in Fig. 4, is used to preform tests for the static and low-speed loading rate ranges (Table 3). The loading rates are calculated assuming engineers’ theory of

Fig. 4—Three-point bending tests for lower three rates of loading (10–6 to 10–4 s–1).

bending and based on Young’s modulus values resulting from compressive strength experimental tests at the quasistatic strain rate. The loading rate of the machine is verified by testing two prisms supported on load cells and the results showed that the machine accurately recorded the force and time. The high-speed dynamic tests have been conducted using a drop-weight impact technique. A small drop-hammer apparatus is designed at Ryerson University to test prisms under higher strain rates (Table 3). The schematic diagram of the setup and the test configuration is illustrated in Fig. 5. The system has the capacity to drop a 37.5 kg (82.7 lb) mass from heights of up to 1200 mm (47 in.). The drop-hammer is solid steel cylinder and it is supported and guided by a steel frame. The striking surface of the drop-hammer is flat circular of 51 mm (2 in.) diameter. In this study, three drop heights— 150, 300, and 600 mm (5.9, 11.8, and 23.4 in.)—are adopted and three specimens are tested at each drop height. The impact force is determined from the average reading of two (±2000g) accelerometers mounted to the drophammer. In addition, the reaction forces between the support and the specimens are measured using dynamic load cells. No damping materials are used in the contact zone between the hammer and the specimen during the tests, as that inadvertently reduces the strain rate. Additionally, all specimens are visually inspected after testing.

Fig. 5—Drop-weight impact test setup for higher three rates of loading. (Note: 1 m = 3.28 ft; 1 mm = 0.0394 in.) Table 3—Summary of flexural strength tests Matrix

Machine Quasi-static MTS 793

HSC

Low-speed range

FRCC1 FRCC2 Drop weight

High-speed loading range

*

Displacement rate for HSC based on elastic modulus of 30 GPa (4382.78 ksi).

†

Drop height in mm (in.).

146

Strain rate, s–1

Loading rate, mm/min (in./min)

1 × 10–6

0.013 (0.0005)*

1 × 10–5

0.130 (0.0051)*

1 × 10

1.30 (0.0510)*

–4

From testing

150 (5.90)†

From testing

300 (11.80)†

From testing

600 (23.60)†

ACI Materials Journal/March-April 2016

Fig. 6—Typical impact and reaction forces versus time. Table 4—Characteristic mechanical properties Matrix

fcʹ,* MPa (ksi)

fr,† MPa (ksi)

Ec,‡ GPa (ksi)

HSC

83.10 (12.05)

8.00 (1.16)

30.22 (4382.78)

FRCC1

110.80 (16.07)

12.10 (1.76)

33.85 (4908.97)

FRCC2

132.70 (19.25)

13.73 (2.00)

39.32 (5702.67)

*

Compressive strength.

†

Flexural strength.

‡

Elastic modulus.

Fig. 7—Stress-strain curves for three tested matrixes (average curves).

This technique is previously calibrated using two drop tests from a 200 mm (8 in.) height direct on a calibrated load cell. To absorb the vibration results from impact and minimize the quantity of noise in the acquired data, a 50 mm (2 in.) fine sand layer is used as a support for the concrete platform. A dynamic data acquisition signal analyzer, with an eight-channel dynamic analyzer, is used for data analysis. The system is provided with 300 MHz sampling integrated electronic piezoelectric (IEPE) sensors capable of data capture, playback, shock recording, analysis, and software processing. Figure 6 shows the impact and the reaction forces versus time curves. Comparing the impact force with the reaction force, it is obvious that the peak load of the impact force is much greater than that of the reaction force. The reason is that most of the impact force is used to balance the inertia force, while a small portion of impact force is used to deform and fracture of specimens.21,22 Thus, the flexural load for each drop is calculated by summing the two support reactions.22 EXPERIMENTAL RESULTS AND DISCUSSION Basic static mechanical properties The characteristic mechanical properties tested at the lowest (quasi-static) strain rate are listed in Table 4. The compressive stress-strain curves are shown in Fig. 7, and the flexural responses of all tested matrixes are illustrated in Fig. 8. It can be first observed that FRCC matrixes show strain hardening behavior under both compression and flexural loading. Additionally, both the strength and the maximum post-cracking strain are significantly improved by using steel fibers, especially under flexural loading. Moreover, the descending branch of FRCC1 and FRCC2 curves has approximately the same slope because both contain 2% fiber by volume.

ACI Materials Journal/March-April 2016

Fig. 8—Flexural responses of three tested matrixes. DIFs of compressive strength and modulus of elasticity Fifty-four cylinders are tested to determine the strain rate effect on compressive strength and elastic modulus. Table 5 summarizes the test results for the three tested matrixes. Each data point in the table is averaged from three specimens, as previously mentioned. It is found that the mechanical properties increase with the increase in the loading rate; all results for different matrixes show the same trend. DIF is much higher 147

Table 5—Compressive strength and elastic modulus experimental results Matrix

Strain rate, s–1

3 × 10–5

3 × 10–4

3 × 10–3

1 × 10–2

3 × 10–2

1 × 10–1

fcʹ, MPa

83.1

85.5

89.4

†

90.8

93.3

94.7

DIF

1.00

1.03

1.08

1.09

1.12

1.14

Ec, GPa

30.2

31.9

34.4

†

35.0

36.7

38.4

DIF‡

1.00

1.06

1.14

1.16

1.21

1.27

fcʹ,* MPa

110.8

112.8

114

117.9

119.5†

120.8

DIF

1.00

1.02

1.03

1.06

1.08

1.09

33.8

34.7

35.5

36.8

37.9

39.8

*

HSC

FRCC1

§

‡

Ec, GPa §

DIF

FRCC2

‡

†

1.00

1.03

1.05

1.09

1.12

1.18

fcʹ,* MPa

132.7

133.9

136.1

137.0

139.1

143.0

DIF‡

1.00

1.01

1.03

1.03

1.05

1.08

Ec, GPa

39.3

40.1

42.1

42.9

45.0

44.9

DIF

1.00

1.02

1.07

1.09

1.14

1.14

‡

§

‡

Compressive strength.

*

Average of two specimens.

†

DIF is dynamic increase factor with respect to static case.

‡

Elastic modulus.

§

Notes: 1 MPa = 145 psi; 1 GPa = 145 ksi.

for matrixes with lower strengths, and the enhancement or DIF is different for compression and elastic modulus. There is significant scatter observed between elastic modulus DIF results of FRCC1 and FRCC2 specimens, especially at higher rates, but it is not thought to be significant. As mentioned previously, the results are fitted with CEB-fib Model Code (Fig. 9). It can be observed that the CEB Model Code gives matching results for the HSC but overestimates both compressive strength and elastic modulus enhancement for FRCC1 and FRCC2 matrixes with compressive strength over 110 MPa (15 ksi). It should be mentioned that the maximum difference between DIF derived from CEB-fib and experimental results in both compressive strength and elastic modulus is less than 6%. DIFs of flexural strength Table 6 shows the flexural strengths for lower three strain rates obtained using the hydraulic servo-controlled testing machine, and results from the dynamic flexural testing program are given in Table 7. Analyzing the tests results, the flexural tensile strength is more sensitive than the compressive strength and elastic modulus at same strain rate. Additionally, DIF is higher for matrixes with lower strengths. As shown in Fig. 10, there is no significant variation in crack pattern observed for different strain rates, even at higher strain rates using drop-weight impact machine. The cracking mode indicated that the specimens are failed in bending (tension side). No compression damage or inclined cracks are observed in any specimens. Additionally, fibers pullout is the only observed mode of failure for FRCC matrixes. Pullout of short straight fibers has previously been shown to be independent of strain rate.13,14 However, the fracture surface became more flattened with the increasing strain rate. This can be explained by the time taken for microcrack propagation in the matrix. At low loading rate, the microcrack grows through the path of lowest strength. 148

Fig. 9—Comparison between DIF derived from tests and CEB-fib for compression. ACI Materials Journal/March-April 2016

However, at higher loading rates, the microcrack will not have time to develop laterally into defected zones and will instead follow a direct path through a stronger zone. The dynamic flexural strength enhancement of the experimental results is shown in Fig. 11. The results are compared with the model for strain rate enhancement of tensile strength of CEB-fib Model Code 2010. It can be seen that the DIF is greatest for specimens without fibers and with lower compressive strength for specimens containing fibers. Addi-

tionally, CEB-fib overestimates tensile enhancement for FRCC1 and FRCC2 matrixes. However, it gives matching results for the HSC matrix. Additionally, the maximum difference between DIF derived from CEB-fib and experimental results is greater than 15%, which is significant if compared with the difference in compressive strength. CONCLUSIONS An experimental investigation was conducted to determine the dynamic behavior of high-strength concrete (HSC) and fiber-reinforced cementitious composites (FRCCs) containing 2% short steel fibers. Compressive strength, modulus of elasticity, and flexural tensile strength has been investigated under six different strain rates, ranging from the static (10–5 s–1) to the seismic and/or impact level (1 s–1). The compressive strengths of the tested matrixes are 83, 110, and 130 MPa (12, 15, and 19 ksi), respectively. A special impact setup was designed at Ryerson University and used to determine the dynamic flexural strength of tested matrixes. The following conclusions can be drawn from the experimental study that was conducted:

Table 6—Rate effect on flexural strength in low-speed loading rates Matrix

HSC

Strain rate, s–1 fr,* MPa

FRCC1

FRCC2

DIF†

fr,* MPa

DIF†

fr,* MPa

DIF†

10

8.0

1.00

12.10

1.00

13.70

1.00

10

––

––

12.40

1.02

13.90

1.01

10

8.95

1.23

12.65

1.05

14.12

1.03

–6 –5 –4

*

Flexural strength.

†

DIF is dynamic increase factor with respect to static case.

Notes: 1 MPa = 145 psi.

Table 7—Rate effect on flexural strength in high-speed loading rates Matrix Drop height, mm

HSC Strain rate, s

–1

FRCC1

fr, MPa

DIF

*

†

Strain rate, s

–1

FRCC2

fr, MPa

DIF

*

†

Strain rate, s

–1

fr,* MPa

DIF†

150

0.32

10.05

1.26

0.85

14.10

1.17

0.79

15.25

1.11

300

1.63

10.45

1.31

1.55

14.25

1.18

1.33

15.40

1.12

600

2.55

10.65

1.33

2.58

14.50

1.20

2.50

15.75

1.15

*

Flexural strength.

†

DIF is dynamic increase factor with respect to static case.

Notes: 1 mm = 0.039 in.; 1 MPa = 145 psi.

Fig. 10—Failure patterns in tested flexural specimens. (Note: 1 mm = 0.0394 in.) ACI Materials Journal/March-April 2016

149

concrete, offshore design, creep, finite element analysis, and structural health monitoring.

ACKNOWLEDGMENTS

This work is financially supported by the Natural Sciences and Engineering Research Council of Canada (NSERC).

REFERENCES

Fig. 11—Comparison between DIF derived from tests and CEB-fib for tensile strength. 1. The compressive strength, elastic modulus, and the flexural tensile strength increase with an increase in strain rates. However, flexural tensile strength is more sensitive than both compressive strength and elastic modulus at that same strain rate. 2. Strain hardening behavior subsists under high strain rates in both loading cases—compression and flexure. 3. DIF is higher for matrixes with lower strengths in both compression and flexure. 4. Quasi-static bending and drop-weight tests had identical failure modes: cracking is observed in the high moment zone and final fracture occurred by fiber pullout in one localized bending crack at the midspan of specimens. However, the fracture surface became more flattened with the increasing strain rate. 5. The CEB-fib Model (2010) fits reasonably well with HSC results in both compression and tension. 6. The CEB-fib Model (2010) overestimates both compressive and tensile strength enhancement for FRCC matrixes with compressive strength greater than 110 MPa (15 ksi). Additionally, the difference between CEB-fib and experimental results is more significant in tension, as fiber contribution is much more effective. Although this experimental program is based on a strain rate range less than a blast loading range, it is recommended that new and more accurate constitutive models for deriving dynamic strength enhancement for FRCC are developed, especially for those of compressive strength greater than CEB-fib limit (120 MPa [17 ksi]), that is, for ultra-highperformance fiber-reinforced concrete with compressive strength greater than 150 MPa (22 ksi). AUTHOR BIOS

H. Othman is a PhD Candidate in the Department of Civil Engineering at Ryerson University, Toronto, ON, Canada. He received his BSc and MSc from Zagazig University, Zagazig, Egypt, and Menoufia University, Al Minufya, Egypt, respectively. His research interests include high-strainrate material response, analysis and modeling of ultra-high-performance fiber-reinforced concrete under dynamic loads, and finite element analysis. H. Marzouk, FACI, is a Professor of the Civil Engineering Department at Ryerson University. He received his MSc and PhD from the University of Saskatchewan, Saskatoon, SK, Canada. He is a member of ACI Committees 209, Creep and Shrinkage in Concrete, and 213, Lightweight Aggregate and Concrete. His research interests include structural and material properties of high-strength and ultra-high-performance fiber-reinforced

150

1. Habel, K., and Gauvreau, P., “Response of Ultra-High Performance Fiber Reinforced Concrete (UHPFRC) to Impact and Static Loading,” Cement and Concrete Composites, V. 30, No. 10, 2008, pp. 938-946. doi: 10.1016/j.cemconcomp.2008.09.001 2. Li, Q.; Reid, S.; Wen, H.; and Telford, A., “Local Impact Effects of Hard Missiles on Concrete Targets,” International Journal of Impact Engineering, V. 32, No. 1-4, 2005, pp. 224-284. doi: 10.1016/j.ijimpeng.2005.04.005 3. Malvar, L., and Ross, C., “Review of Strain Rate Effects for Concrete in Tension,” ACI Materials Journal, V. 95, No. 6, Nov.-Dec. 1998, pp. 735-739. 4. Bischoff, P., and Perry, S., “Compressive Behaviour of Concrete at High Strain Rates,” Materials and Structures, V. 24, No. 6, 1991, pp. 425-450. doi: 10.1007/BF02472016 5. Williams, M., “Modeling of Local Impact Effects on Plain and Reinforced Concrete,” ACI Structural Journal, V. 91, No. 2, Mar.-Apr. 1994, pp. 178-187. 6. Ross, C.; Tedesco, J.; and Kuennen, S., “Effects of Strain Rate on Concrete Strength,” ACI Materials Journal, V. 92, No. 1, Jan.-Feb. 1995, pp. 37-47. 7. Li, Z., and Huang, Y., “Effect of Strain Rate on the Compressive Strength Surface Cracking and Failure Mode of Mortar,” ACI Materials Journal, V. 95, No. 5, Sept.-Oct. 1998, pp. 512-518. 8. Millard, S.; Molyneaux, T.; Barnett, S.; and Gao, X., “Dynamic Enhancement of Blast-Resistant Ultra-High Performance Fibre-Reinforced Concrete under Flexural and Shear Loading,” International Journal of Impact Engineering, V. 37, No. 4, 2010, pp. 405-413. doi: 10.1016/j. ijimpeng.2009.09.004 9. Wille, K.; Kim, D.; and Naaman, A., “Strain-Hardening UHP-FRC with Low Fiber Contents,” Materials and Structures, V. 44, No. 3, 2011, pp. 583-598. doi: 10.1617/s11527-010-9650-4 10. Banthia, N.; Mindess, S.; and Trc, J., “Impact Resistance of Steel Fiber Reinforced Concrete,” ACI Materials Journal, V. 93, No. 5, Sept.-Oct. 1996, pp. 472-479. 11. Gopalaratnam, V., and Shah, S., “Properties of Steel Fiber Reinforced Concrete Subjected to Impact Loading,” ACI Journal Proceedings, V. 83, No. 4, July-Aug. 1986, pp. 117-126. 12. Maalej, M.; Quek, S.; and Zhang, J., “Behavior of Hybrid-Fiber Engineered Cementitious Composites Subjected to Dynamic Tensile Loading and Projectile Impact,” Journal of Materials in Civil Engineering, ASCE, V. 17, No. 2, 2005, pp. 143-152. doi: 10.1061/(ASCE)0899-1561(2005)17:2(143) 13. Gokoz, U., and Naaman, A., “Effect of Strain-Rate on the Put-Out Behaviour of Fibers in Mortar,” International Journal of Cement Composites and Lightweight Concrete, V. 3, No. 3, 1981, pp. 187-202. doi: 10.1016/0262-5075(81)90051-8 14. Suaris, W., and Shah, S., “Strain-Rate Effects in Fibre-Reinforced Concrete Subjected to Impact and Impulsive Loading,” Composites, V. 13, No. 2, 1982, pp. 153-159. doi: 10.1016/0010-4361(82)90052-0 15. Comité Euro-International du Béton, “CEB-fib Model Code 2010,” 1st Draft, Volume 1, Lausanne, Switzerland, 2010. 16. Marzouk, H., “Creep of High-Strength Concrete and NormalStrength Concrete,” Magazine of Concrete Research, V. 43, No. 155, 1991, pp. 121-126. doi: 10.1680/macr.1991.43.155.121 17. Rossi, P.; Arca, A.; Parant, E.; and Fakhri, P., “Bending and Compressive Behaviours of a New Cement Composite,” Cement and Concrete Research, V. 35, No. 1, 2005, pp. 27-33. doi: 10.1016/j.cemconres.2004.05.043 18. Acker, P., and Behloul, M., “Ductal® Technology: A Large Spectrum of Properties, A Wide Range of Applications,” International Symposium on Ultra High Performance Concrete, 2004, pp. 11-24 19. Yazdizadeh, Z., “Use of Fiber Brag Gating Sensors in Civil Engineering Applications,” MSc thesis, Ryerson University, Toronto, ON, Canada, 2014, pp. 49-79. 20. Comité Euro-International du Béton, “Concrete Structures under Impact and Impulsive Loading,” Synthesis Report, CEB Bulletin No. 187, Lausanne, Switzerland, 1988, 184 pp. 21. Soleimani, S., and Banthia, N., “A Novel Drop Weight Impact Setup for Testing Reinforced Concrete Beams,” Experimental Techniques, V. 38, No. 3, 2014, pp. 72-79. doi: 10.1111/j.1747-1567.2012.00810.x 22. Zhang, X.; Ruiz, G.; Yu, R.; and Tarifa, M., “Fracture Behaviour of High-Strength Concrete at a Wide Range of Loading Rates,” International Journal of Impact Engineering, V. 36, No. 10-11, 2009, pp. 1204-1209. doi: 10.1016/j.ijimpeng.2009.04.007

ACI Materials Journal/March-April 2016

ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M15

Analysis of Compressive Strength Development and Carbonation Depth of High-Volume Fly Ash Cement Pastes by Xiao-Yong Wang and Ki-Bong Park High-volume fly ash (HVFA) concrete, which typically has 50 to 60% fly ash as the total cementitious material content, is widely used to achieve sustainable development in the concrete industry. Strength development and carbonation are critical research topics for using HVFA concrete. This paper presents a numerical procedure to evaluate the strength development and carbonation depth of HVFA concrete. This numerical procedure consists of a hydration model and a carbonation reaction model. The hydration model analyzes the fly ash dilution effect and the pozzolanic reaction. The amount of carbonatable materials, such as calcium hydroxide (CH) and calcium silicate hydrate (CSH), are calculated using reaction degrees of cement and fly ash. The compressive strength development of cement-fly ash blends are evaluated from CSH contents. The calculation results from the hydration model, such as the amount of carbonatable materials and the porosity, are used as input parameters for the carbonation reaction model. By considering the effects of material properties and environmental conditions, the carbonation reaction model analyzes the diffusivity of carbon dioxide and the carbonation depth of HVFA concrete with different curing conditions, different fly ash contents, and different water-binder (w/b) ratios. Keywords: carbonation; compressive strength; dilution effect; high-volume fly ash; hydration; model; pozzolanic reaction.

INTRODUCTION Fly ash consists of finely divided ashes produced by burning pulverized coal in power stations and can be categorized as a normal type of pozzolan to produce high-strength and high-performance concrete. To achieve sustainable development in the concrete industry, high-volume fly ash (HVFA) concrete, which typically has 50 to 60% fly ash as the total cementitious material content, is widely used. The incorporation of a high volume of fly ash in concrete has many advantages such as reducing water demand, improving workability, minimizing cracking due to thermal and drying shrinkage, and enhancing durability to sulfate attack and alkali-silica expansion.1 Compressive strength is the most important property of hardened concrete; other properties, such as mechanical properties development and construction management, are closely related to compressive strength development. However, due to carbonation in reinforced concrete structures, when the pH of the capillary pore water drops to a low value of 9, the passive layer on the steel surface will no longer remain stable and corrosion of the steel reinforcing bar will begin. Therefore, the compressive strength development and carbonation are critical research topics for materials selection, durability design, and maintenance of reinforced concrete structures.1

ACI Materials Journal/March-April 2016

Many experimental studies have examined the strength development and carbonation of HVFA concrete. Lam et al.2 found that fly ash contributed little to compressive strength at early ages, and at later ages, the contribution of fly ash to the compressive strength became larger. The contribution of fly ash in concrete mixtures prepared at a lower water-cementitious materials ratio (w/cm) was greater than those prepared at a higher w/cm. Papadakis3 and Papadakis et al.4,5 found that for fly ash blended concrete, the carbonation depth decreases as aggregate replacement by fly ash increases and also increases as cement replacement by fly ash increases. Sisomphon and Franke6 and Jiang et al.7 found that an effective water-binder ratio (w/b) and cement content are the key factors affecting HVFA concrete carbonation. An increased curing period can improve the carbonation behavior of HVFA concrete.6,7 According to References 2 through 7, the strength and carbonation of HVFA concrete is closely related to the material properties of concrete, such as w/b, fly ash replacement ratios, and curing periods. Compared to the abundant experimental studies, theoretical models for evaluating the strength development and carbonation of HVFA concrete are limited. Using an apparent activation energy function, Han et al.8 and Kim et al.9 evaluated the development of the compressive strength of hardening fly ash blended concrete, investigating the influences of fly ash replacement content and the w/b on the apparent activation energy. Based on experimental results concerning the compressive strength development of concrete containing fly ash, Hwang et al.10 derived an estimation equation for the compressive strength development. The equation used a coefficient to indicate the activity of fly ash as a binder in the form of a function of age, fly ash content, and the Blaine specific surface area of fly ash. Conversely, Papadakis3 proposed a simplified scheme describing the activity of fly ash in terms of chemical reactions and yielded quantitative expressions of the final chemical composition of supplementary cementitious materials (SCM) concrete. The carbonation depth of concrete incorporating low-volume fly ash (fly ash content of less than 30% of the total binder content) was predicted considering both material properties and exposed conditions. Using an effective w/b, Jiang et al.7 modified Papadakis’ original equation3 and predicted the carbonation of HVFA concrete. However, Papadakis3 and Jiang et al.7 focused on matured concrete. The depenACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-056.R2, doi: 10.14359/51688636, received July 8, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

151

Table 1—Comparison between proposed model and previous models Strength development evaluation of HVFA blended concrete

Carbonation depth evaluation of HVFA blended concrete

Papadakis3

No

Yes

7

Jiang et al.

No

Yes

Han et al. ; Kim et al.9

Yes

No

Hwang et al.10

Yes

No

Proposed model

Yes

Yes

8

dence of carbonation resistance on the curing period was not considered in detail in previous research.3,7 Summarily, current models3,7-10 are only valid for single-property evaluation of fly ash blended concrete, such as strength development evaluation or carbonation evaluation. An integrated model that can evaluate both compressive strength development and carbonation is necessary. To overcome the weak points in former research,3,7-10 this paper presents a numerical procedure to evaluate the strength development and carbonation depth of HVFA concrete. The comparison among our proposed model and previous models3,7-10 is shown in Table 1. Due to the combination of the blended cement hydration model with the carbonation reaction model, the proposed model shows more functions than previous models.3,7-10 The flowchart of the numerical procedure is shown in Fig. 1. Using a hydration model considering both cement hydration and fly ash reaction, the amounts of calcium hydroxide (CH), chemically bound water, and calcium silicate hydrate (CSH) are determined as functions of the curing age. The compressive strength development of cement-fly ash blends are evaluated from CSH contents. Furthermore, by considering the effects of material properties and environmental conditions, the diffusivity of carbon dioxide and the carbonation depth of concrete are calculated. RESEARCH SIGNIFICANCE Compressive strength development and carbonation are critical research topics for using HVFA concrete. Using an HVFA blended hydration model, this paper analyzes the compressive strength development of concrete through the reaction degrees of cement and fly ash. By combining the hydration model with the carbonation reaction model, the effects of w/b, fly ash replacement ratios, and curing periods on the carbonation resistance of HVFA concrete are detailed clearly. The proposed numerical procedure is useful for carbonation durability design and mixing proportions selection for HVFA concrete. HYDRATION MODEL FOR CEMENT-FLY ASH BLENDS Hydration model of portland cement Park et al.11 proposed a shrinking-core model to model the hydration of portland cement. This model is expressed

152

Fig. 1—Flowchart of numerical procedure. as a single equation consisting of three coefficients: kd, the reaction coefficient in the induction period; De, the effective diffusion coefficient of water through the CSH gel; and kri, a coefficient of the reaction rate of the mineral compound of cement, as shown in Eq. (1) and (2) d α i 3 ( S w / S0 ) ρw Cw-free = × dt (v + wg )r0 ρc

1

−2

(1)

−1  1 r0  r0 1 3 + ( 1 − α ) + (1 − α i ) 3 − i  k  D D k d e e ri 4



α=

∑ α i gi

i =1 4

∑ gi

(2)

i =1

where αi (i = 1, 2, 3, and 4) represents the reaction degree of the mineral compound of cement C3S, C2S, C3A, and C4AF, respectively; α is the degree of cement hydration and can be calculated from the weight fraction of the mineral compound gi and the reaction degree of the mineral compound αi; ν is the stoichiometric ratio by mass of water to cement (= 0.25); wg is the physically bound water in the CSH gel (= 0.15); ρw is the density of water; ρc is the density of cement; Cw-free is the amount of water at the exterior of the CSH gel; r0 is the radius of unhydrated cement particles; Sw is the effective surface area of the cement particles in contact with water; and S0 is the total surface area if the surface area develops unconstrained. The reaction coefficient kd is assumed to be a function of the degree of hydration, as shown in Eq. (3), where B and C are the coefficients determining this factor; B controls the rate of the initial shell formation and C controls the rate of the initial shell decay.

kd =

B + Cα 3 (3) α1.5

ACI Materials Journal/March-April 2016

Table 2—Coefficients of cement hydration model B20, cm/h 8.09 × 10

–9

C20, cm/h 0.02

krC3S 20, cm/h

krC2S 20, cm/h

krC3 A20, cm/h

9.03 × 10

2.71 × 10

1.35 × 10

–6

–7

–6

krC4 AF 20, cm/h 6.77 × 10

–8

De20, cm2/h

β1, K

β2, K

β3, K

E/R, K

8.62 × 10

1000

1000

7500

5400

–10

Notes: 1 cm = 0.394 in.; 1 cm = 0.155 in. ; °F = (K – 273.15) × 1.8 + 32. 2

2

The effective diffusion coefficient of water is affected by the tortuosity of the gel pores and the radii of the gel pores in the hydrate. This phenomenon can be described as a function of the degree of hydration and is expressed as follows

 1 De = De 0 ln   (4)  α

where De0 is initial diffusion coefficient. The amount of water in the capillary pores Cw-free is expressed as a function of the degree of hydration in the previous step, as shown in Eq. (5) r



 W − 0.4 ⋅ α ⋅ C0  Cw-free =  0  (5) W0 

where C0 and W0 are the mass fractions of cement and water in the mixture proportion, respectively; and r is an empirical parameter considering the accessibility of water into an inner anhydrous part through an outer hard shell of the cement particles. (When the w/b is higher than 0.4, r = 1.0; when the w/b is lower than 0.4, because of increased constrictivity and tortuosity of the capillary pore network and less pore connectivity, r is higher than 1 and can be determined from r = 2.6 – 4(W0/[C0 + P]), where P is the mass of mineral mixtures.12,13) The effect of temperature on the reaction coefficients is assumed to follow Arrhenius’s law, as shown in Eq. (6) through (9)

1   1 B = B20 exp  −β1  −  (6)   T 293  



1   1 C = C20 exp  −β 2  −  (7)   T 293  



1   E1 kri = kri 20 exp  −  − (8)   R T 293  



1   1 De = De 20 exp  −β3  −  (9)   T 293  

where β1, β2, E/R, and β3 are the temperature sensitivity coefficients; and B20, C20, kri20, and De20 are the values of B, C, kri, and De at 20°C (68°F), respectively. Based on the degree of reactions of mineral compounds of cement,14 using a predictor-corrector algorithm, the parameters of the hydration model are calibrated and shown in Table 2. This predictor-corrector algorithm proceeds in two steps: first, the prediction step assumes the rough approximations of the desired quantities. Second, the corrector step ACI Materials Journal/March-April 2016

refines the initial approximations using an iteration method. For example, if refining the value of kri, the old values of B, C, and De obtained from last step should be used and the new value of kri can be confirmed according to the experimental results during the activated chemical reaction period. The refined value of kri will be used as an input parameter in the next calculation step. Similarly, the value of De can be refined through the experimental results during the diffusion period, the value of B can be refined through the experimental results on the formation of the initial impermeable layer, and the value of C can be refined through the experimental results on the destruction of the initial impermeable layer. In the next iteration step, the refined values of the coefficients are used as input values for calibration.11 Once the convergence criteria are met, the refining process will stop. Using the proposed portland cement hydration model, Park et al.11 evaluated the heat evolution rate, chemically bound water, and compressive strength of hardening concrete. However, Park’s model is only valid for portland cement concrete. To evaluate the properties of HVFA concrete, the fly ash reaction and the interactions between cement hydration and the fly ash reaction should be considered. Simulation of the pozzolanic reaction in cementfly ash blends The hydration rate of pozzolanic materials depends on the amount of CH in hydrating cement-fly ash blends and the reaction degree of the mineral admixtures.15-19 Compared to silica fume, the hydration rate of fly ash is much lower due to the larger particle size. The simulation assumes that the reaction of the fly ash is divided into three processes: an initial dormant period, a phase-boundary reaction process and a diffusion process. By considering these points, the reaction equation of the fly ash is originally proposed as follows d α FA mCH (t ) 3ρw = × dt P vFA rFA0ρFA

1 −1 −2  1  rFA0 rFA0 1 3 3  k − D  + D (1 − α FA ) + k (1 − α FA ) eFA rFA dFA eFA (10)





kdFA =

BFA + CFA ⋅ (α FA )3 (11) (α FA )1.5

 1  DeFA = DeFA0 ⋅ ln  (12)  α FA 

where αFA is the reaction degree of the fly ash; P is the mass of the fly ash in the mixture proportion; mCH(t) is the CH mass in a unit volume of hydrating cement-fly 153

ash blends; νFA is the stoichiometric ratio of fly ash to CH (νFA = 0.845 – 0.7(P/[C0 + P])[17]); rFA0 is the radius of the fly ash particle; ρFA is the density of the fly ash; kdFA is the reaction rate coefficient in the dormant period (BFA and CFA are coefficients); DeFA0 is the initial diffusion coefficient; and krFA is the reaction rate coefficient. Similar to the cement hydration model, the influence of the temperature on the fly ash reaction can be considered using the Arrhenius law.15 Mutual interaction between cement hydration and fly ash reaction In the model, the amount of free water left in the system and the amount of CH were adopted as the control indicators for the reactions of cement-fly ash blends. The amounts of CH, chemically bound water, and capillary water in cement-fly ash blends during hydration can be determined with the following equations

mCH(t) = RCHCE ∙ C0 ∙ α – νFA ∙ αFA ∙ P (13)

hardening fly ash blended concrete. Equation (17) assumes that the chemical compositions of fly ash, such as SiO2 and Al2O3, react at the same rate. Using the blended hydration model, the phase volume fractions of hydrating cement-fly ash paste can be calculated as follows







Wcap = W0 – 0.4 ∙ C0 ∙ α – RCWFA ∙ αFA ∙ P – RPWFA ∙ αFA ∙ P (14)

Wcbm = ν ∙ C0 ∙ α + RCWFA ∙ αFA ∙ P (15)

In Eq. (13) through (15), mCH(t), Wcap, and Wcbm are the masses of CH, capillary water, and chemically bound water, respectively; RCHCE is the mass of produced CH from 1 g (0.0022 lb) of hydrated cement; RCWFA is the mass of chemically bound water from 1 g (0.0022 lb) of reacted fly ash (RCWFA = 0.1[15]); and RPWFA is the mass of gel water from 1 g of reacted fly ash (RPWFA = 0.15[15]). As shown in Eq. (13) through (15), the evolution of CH, chemically bound water, and capillary water in cement-fly ash blends depends on both cement hydration and fly ash reaction. Papadakis and Tsimas18,19 and Papadakis20 proposed that for matured fly ash blended concrete, the CSH content, which is the most critical parameter in strength development, can be calculated as a function of the cement content C0; fly ash content P; weight fraction of SiO2 in cement fS,C and fly ash fS,P, respectively; and ratio of active silica to total silica in the fly ash γs. The original chemical reaction equation proposed by Papadakis and Tsimas18,19 and Papadakis20 is as follows

CSH = 2.85 (fS,C ∙ C0 + fS,P ∙ P ∙ γs)

(16)

where the coefficient 2.85 is the ratio between the molar weight of CSH and the weight of oxide SiO2 in CSH.18,19 Equation (16) is only valid for matured concrete. To evaluate the strength development of hardening concrete, Eq. (16) was combined with with the blended cement hydration model. The revised equation is as follows CSH(t) = 2.85(fS,C ∙ C0 ∙ α + fS,P ∙ P ∙ αFA)

(17)

In Eq. (17), the upper limit of the reaction degree of cement α is 1, and the upper limit of the reaction degree of fly ash αFA is γs. Hence, Eq. (16) can be regarded as the upper limit of the revised Eq. (17). By combining Eq. (16) with the reaction degrees of binders, the revised Eq. (17) is valid for 154



V1 = V2 =

C0 (1 − α ) (18) ρc

P (1 − α FA ) (19) ρFA

V3 = V4 =

CSH(t ) (20) ρCSH

W0 − Wcbm ρw

(21)

V5 = 1 – V1 – V2 – V3 – V4 (22)

where V1, V2, V3, V4, and V5 are the volumes of unhydrous cement, unreacted fly ash, CSH (ρCSH is density of CSH3), porosity, and other hydration products, respectively. The addition of fly ash mainly represents the dilution effect and the chemical effect on cement hydration.21,22 The dilution effect is a consequence of the replacement of cement by fly ash and increases the w/c. This dilution effect is considered through the amount of capillary water (Eq. (14)) and the dilution effect (Eq. (5)). The chemical effect is the pozzolanic reaction between fly ash and CH and is considered using Eq. (10) through (12). Conversely, the addition of fly ash also can retard the hydration of cement in the early ages. This retardation effect comes from the dissolution of aluminate ions or organic matter from the fly ash.18-22 In addition, fly ash particles can serve as nucleation sites for the cement particles and accelerate the hydration of cement. Because of the coexistence of both the retardation effect and the acceleration effect, as reported by Papadakis and Tsimas18,19 and Papadakis,20 when the aggregate is partly replaced by fly ash, the early-age compressive strength and CH amount are not significantly different from those of the control portland cement concrete. Therefore, the hydration coefficients of cement in cement-fly ash blends are assumed to be approximately the same as those in plain portland cement concrete. Evaluating properties of cement-fly ash blends Evaluating reaction degree of fly ash—Lam et al.17 measured the reaction degree of fly ash in cement-fly ash paste with different w/b and fly ash replacement ratios. Cement-fly ash pastes were prepared at w/b of 0.19, 0.24, 0.3, and 0.5. Fly ash was used to replace cement at levels of 25% and 45% by weight for the pastes at w/b of 0.19 and 0.24, respectively, and 25% and 55% by weight for those at w/b of 0.3 and 0.5, respectively. Plain portland cement pastes without any fly ash replacement were prepared at the same w/b as the references. The paste specimens were cured ACI Materials Journal/March-April 2016

Fig. 2—Reaction degree of fly ash. in water at 27°C (80.6°F). At the ages of 7, 28, and 90 days, the degree of reaction of fly ash was measured based on a selective dissolution procedure using a picric acid-methanol solution and water. The CH contents were measured using thermal gravimetry analysis, the chemically bound water contents were measured using loss on ignition in an electric furnace, and the compressive strength test was performed using a hydraulic compression machine. For cement-fly ash paste, the development of properties relate to both the cement hydration and the fly ash reaction. The contribution from the cement hydration can be evaluated using the coefficients of the cement hydration model (obtained in the section titled “Hydration model of portland cement”). From the experimental results of the reaction degree of fly ash, using the predictor-corrector algorithm, the coefficients relating to the fly ash reaction can be calibrated and are shown in Table 3. As shown Fig. 2, the analyzed results for the reaction degree of fly ash generally agree with the experimental results. Reducing the replacement level of the fly ash increases both the alkaline activating effect of the cement and the reactivity of fly ash. Increasing the w/b creates more available space for hydration products to form, and the reactivity of the fly ash increases correspondingly. Because the interactions between cement hydration and fly ash reaction are considered, the reaction coefficients of fly ash do not change with the w/b or fly ash replacement ratios. Figure 3 presents the calculation results of CH in cement-fly ash blends. For portland cement paste, the amount of CH will increase until it reaches steady state. For cement-fly ash paste, the evolution of CH depends on two factors: the portACI Materials Journal/March-April 2016

Table 3—Coefficients of fly ash reaction model BFA, cm/h

CFA, cm/h

krFA, cm/h

DeFA0, cm2/h

2.6 × 10–11

0.53

7.21 × 10–7

7.05 × 10–13

Notes: 1 cm = 0.394 in.; 1 cm2 = 0.155 in.2

land cement hydration that produces CH and the pozzolanic reaction that consumes CH. In the initial 7 days, the hydration of portland cement is dominant, so CH will increase continuously and present a peak value at the age of approximately 1 week. After this age, the pozzolanic reaction of fly ash will become dominant, so the CH will decrease. Figure 4 presents the calculation results for chemically bound water in cement-fly ash blends. As shown in Fig. 4(a), for cement-fly ash with a higher w/b of 0.5, when 25% of the cement is replaced with fly ash, the amount of chemically bound water significantly decreases. Figure 4(b) shows that for cement-fly ash with a lower w/b of 0.19, when 25% of the cement is replaced with fly ash, the amount of chemically bound water is comparable to that of portland cement paste due to the dilution effect from the addition of fly ash. Increasing the amount of mineral admixtures decreases the amount of cement, and, consequently, increases the water to cement ratio and increases the degree of hydration of cement. This dilution effect is considered by the cement hydration model in Eq. (5) and is shown in Fig. 5. Figure 5(b) shows that when the w/b is lower, the dilution effect will become more significant. Figure 6 presents the evolution of the phase volume fractions of hardening cement-fly ash blends paste (w/b of 0.4 with 25% fly ash). As shown in this figure, as the cement 155

Fig. 3—Calcium hydroxide contents. (Note: 1 g = 0.0022 lb.)

Fig. 4—Chemically bound water contents. (Note: 1 g = 0.0022 lb.)

Fig. 5—Effect of fly ash addition on hydration of cement. hydration and fly ash reaction proceed, the volumes of unreacted cement and fly ash decrease, the volumes of CSH and other reaction products increase, and, due to the filling effects of reaction products, the pore volume decreases. At an early age, the cement hydration and fly ash reaction proceed quickly, and at a later age, the reaction rates become slower. Because the reactivity of cement is much higher than that of fly ash, at the age of 180 days, the remaining unhydrous cement is much less than that of fly ash. Evaluating compressive strength of cement-fly ash paste— The compressive strength of concrete is closely related to the w/b and the fly ash content. The relation among compres-

156

sive strength, w/b, and fly ash content can be described as follows21

f c (t ) = A1 (t ) ⋅

C0 + k (t ) ⋅ P − A2 (t ) (23) W0

where fc is the compressive strength of concrete; A1(t) and A2(t) are strength coefficients; and k(t) is the efficiency factor of fly ash. In Eq. (23), the mass of binder C0 + k(t) ∙ P in the numerator relates to the mass of reaction products that contribute to the compressive strength; the mass of water W0 in the denominator relates to the available pore space in which hydration products form. However, Eq. (23) has some limits. For hardACI Materials Journal/March-April 2016

Fig. 6—Phase volume fractions of cement-fly ash paste (w/b of 0.4 with 25% fly ash). ening concrete, the coefficients k(t), A1(t), and A2(t) are not constants but are age-dependent variables. With changing w/b, fly ash replacement ratios, and curing ages, the coefficients k, A1, and A2 are different. Due to variances of coefficients, it is not convenient to use Eq. (23) for evaluating the compressive strength development of fly ash blended concrete. In this research, to overcome the weak points of the current model (Eq. (23)), we proposed that the compressive strength of concrete can be determined from CSH contents. The relation between the compressive strength of concrete and the CSH contents can be described using a linear equation as follows

f c (t ) = A1 ⋅

CSH(t ) − A2 (24) W0

In Eq. (24), the mass of calcium silicate hydrate CSH(t) can be determined from Eq. (17). CSH(t) relates to the w/b, the fly ash replacement ratios, and the curing age of concrete. Because the effects of mixing proportions and curing age have been included in the CSH(t) item in Eq. (24), the coefficients of A1 and A2 are constants, not age-dependent variables. As shown in Eq. (24), for hardening concrete, the compressive strength development starts after a threshold degree of hydration. When the degree of hydration is lower than this threshold degree of hydration, the compressive strength of concrete is zero.21,22 The concept of this threshold degree of hydration is similar to that of the final setting time of concrete. (Final set means complete solidification and beginning of hardening. In concrete technology, the phenomenon of strength gain with time is called hardening.21,22) Based on the calculated CSH contents and the measured compressive strength of paste, the strength coefficients of Eq. (24) can be calibrated and are shown in Fig. 7. The value of A1 is given as 97.94 MPa (14.20 ksi), and the value of A2 is 35.53 MPa (5.15 ksi). The correlation coefficient between the experimental results and the predicted results is 0.95. The differences between the experimental results and the analyzed results mainly come from the ignorance of the ACI Materials Journal/March-April 2016

Fig. 7—Compressive strength versus CSH contents. (Note: 1 g = 0.0022 lb; 1 MPa = 145 psi.) CSH distributions in the pore space. The reaction products distribute more homogeneously in pastes with lower w/b than those with higher w/b.21,22 Figure 8 shows the analysis results for compressive strength development of cement-fly ash paste. At a late age, for concrete with HVFA, a relatively higher w/b marginally increases the reaction degree of cement (shown in Fig. 5(a)), so the compressive strength of HVFA paste cannot surpass that of the control paste (Fig. 8(a) with a w/b of 0.5). When the w/b is lower, due to the significant increase in the reaction degree of cement (shown in Fig. 5(b)), the compressive strength of HVFA paste can surpass that of the control paste (Fig. 8(c) with a w/b of 0.24; and Fig. 8(d) with a w/b of 0.19). Therefore, the efficiency factor of fly ash is dependent on the w/b and is not a constant. The contribution of fly ash mixtures prepared at a lower w/b was greater than those prepared at a higher w/b. However, the proposed model for strength development has some limitations due to ignorance of the aggregate influence. At the macroscopic level, concrete is a composite material consisting of discrete aggregates dispersed in a continuous cement paste matrix. The bonding region or interfacial transition zone (ITZ) in concrete between the matrix and the aggregate is a critical component of the mechanical performance.21 For ordinary- and low-strength concrete, the ITZ is the weak link of concrete, and the compressive strength of concrete is mainly dependent on the strength of the ITZ. Alternately, for high-strength concrete, the strength of concrete relates to the three phases of concrete: the ITZ phase, the bulk paste matrix phase, and the aggregate phase. Therefore, the current model is not perfect and requires improvement to consider more influencing factors for concrete strength development. EVALUATION OF CARBONATION DEPTH OF HVFA BLENDED CONCRETE The carbonation of concrete occurs in the cement paste component of concrete in which the aggregates that constitute the majority of the mass and volume of concrete are essential inert fillers, as far as a certain carbonation is 157

Fig. 8—Compressive strength of cement-fly ash blends. (Note: 1 g = 0.0022 lb; 1 MPa = 145 psi.) concerned. The hydration products of the CH and CSH that are susceptible to carbonation typically constitute 85% of the weight of the mass of hardened cement pastes. The carbonation reactions between CO2 and carbonatable constituents are shown as follows3 CH Ca(OH) 2 + CO 2 K → CaCO3 + H 2 O (25)



CSH (3CaO ⋅ 2SiO 2 ⋅ 3H 2 O) + 3CO 2 K → 3CaCO3 ⋅ 2SiO 2 ⋅ 3H 2 O

(26) Concrete carbonation is a complicated physicochemical process. The process includes the diffusion of atmospheric CO2 into the concrete pores, its dissolution in the aqueous film of these pores, the dissolution of solid Ca(OH)2 in the water of the pores, the diffusion of dissolved Ca(OH)2 in pore water, its reaction with dissolved CO2, and the reaction of CO2 with CSH. In addition, there is a parallel process that includes the hydration of cementitious materials and the reduction of concrete porosity. Papadakis3 and Papadakis et al.4,5 developed and experimentally verified a fundamental and comprehensive reaction model of concrete carbonation. When all of the hydration reaction rates are set to zero (when carbonation experiments were conducted with fully hydrated samples), the simplified model equations can be written as shown in Eq. (27) through (29)

∂[CO 2 ]  ∂  DC = [CO 2 ]( K CH [Ca(OH) 2 ] + 3K CSH [CSH])   ∂x ∂x  (27)

158



∂ [Ca(OH) 2 ] = − K CH [CO 2 ][Ca(OH) 2 ] (28) ∂t



∂ [CSH] = − K CSH [CO 2 ][CSH] (29) ∂t

where DC is the effective diffusivity of CO2; [CO2] is the molar concentration of CO2; KCH and KCSH are the carbonation rate constants of Ca(OH)2 and CSH, respectively; and [Ca(OH)2)] and [CSH] are the molar concentrations of Ca(OH)2 and CSH, respectively. This mathematical model is based on the mass balance of gaseous CO2, solid and dissolved Ca(OH)2, and CSH and accounts for the diffusion and consumption of these substances. In the given initial and boundary conditions, the differential equations can be solved using a finite differential method or finite element method numerically. For the typical range of parameters (especially for relative humidity higher than 55%, where CO2 diffusion controls the carbonation process3), a carbonation front will form that divides the concrete into two different regions: a fully carbonated region and one in which the carbonation has not started at all. The distance between this front and the outer concrete surface is called the carbonation depth, and for the most common one-dimensional cases, its evolution with time is given by a simple analytical expression in terms of the composition and the environmental conditions. The evolution of the carbonation depth xc (in m) with time t (in seconds) is given by the analytical expression as shown in Eq. (30) and (31)3

ACI Materials Journal/March-April 2016

Table 4—Mixture proportions of concrete7 Mixture No.

Cement, kg/m3

Fly ash, kg/m3

Fly ash replacement ratio

w/b

Water-reducing agent, %

Slump, mm

C1

222

0

0

0.6

0.2

55

F1-55

100

122

0.55

0.46

1.7

50

C2

333

0

0

0.45

0.2

75

F2-55

150

183

0.55

0.38

1.7

70

Notes: 1 kg/m = 1.68 lb/yd ; 1 mm = 0.0394 in. 3





3

xc =

2 DC ([CO 2 ]0 / 100)t (30) 0.33CH + 0.214CSH

    RH  2.2  εC DC = A   1 −  (31)  C0 + P + W0   100   ρ ρFA ρw  c

where [CO2]0 is the CO2 content in the ambient air at the concrete surface; εC is the porosity of carbonated concrete and can be determined using the proposed model by Papadakis3; and RH is the ambient relative humidity. A and a are parameters that will be regressed from measured carbonation depths. Based on the proposed hydration model, the amount of CH, CSH, and porosity can be obtained as associated results during the hydration period of cement-FA blended concrete. Furthermore, the carbonation depth can be predicted by using Eq. (30) and (31). Jiang et al.7 investigated the carbonation depth of HVFA blended concrete. The mixing proportions of concrete are shown in Table 4.7 The w/b vary between 0.6 and 0.38 and the fly ash replacement ratio is 55%. The fine aggregate was natural sand, and the coarse aggregate was a crushed limestone, with a maximum size of 31.5 mm (1.24 in.).7 The test specimens for the accelerated test were cured at 25°C (77°F) until the time of acceleration carbonation testing (with two types of curing periods: 28 days wet curing and 90 days wet curing). The accelerated carbonation test was conducted in a test chamber kept at a temperature of 20°C (68°F), a relative humidity of 70%, and a CO2 concentration of 20%. The test specimens were 100 x 100 x 200 mm (0.328 x 0.328 x 0.656 ft). The depth of carbonation was determined by removing a slice approximately 50 mm (0.164 ft) thick from the end of the specimen, spraying the freshly broken samples with a phenolphthalein indicator, and measuring the depth to the color change. Using the difference scheme of Eq. (29), the carbonation depth of the specimens can be calculated. The comparison between the prediction results and the experiment results is shown in Fig. 9 (A = 1.52–6, a = 1.8). The prediction results generally reproduced the experimental results. As shown in Fig. 9(b) and (d), for concrete with the same fly ash replacement ratio of 55%, decreasing the w/b from 0.46 (Fig. 9(b)) to 0.38 (Fig. 9(d) decreased the carbonation depth. As shown in Fig. 9(b) and (c), compared to portland cement concrete (Fig. 9(c)), the incorporation of HVFA into concrete (Fig. 9(b)) increases the carbonation depth. When the initial curing periods increase from 28 days to 90 days, the amount of carbonatable constituents will increase, the porosity will decrease, and the carbonation depth will decrease correspondingly. Figure 9(e) presents an ACI Materials Journal/March-April 2016

integral comparison between the predicted and experimental carbonation depth. The correlation coefficient between them is 0.85, and the root-mean-square error (RMSE) is 2.8 mm. However, the original carbonation model3-5 does not explicitly consider the effect of curing period on the amount of carbonatable materials and carbonation depth. In contrast, due to combining the carbonation model with the hydration model, the proposed numerical procedure in this paper can consider more influencing factors for concrete carbonation than the original model.3-5 The proposed carbonation reaction model considers concrete material properties and environmental conditions on the carbonation depth of concrete. Material properties, including cement and fly ash contents, water contents, porosity, reaction degree of binders, and carbonatable materials contents, and environmental conditions, such as relative humidity and carbon dioxide concentration, are considered in this modeling. However, for concrete at early ages, the presence of aggregate appears to influence the carbonation depth results.5,23 This point is not considered in the current proposed model, and more improvements are necessary. CONCLUSIONS This paper presents a numerical procedure to evaluate the strength development and carbonation depth of HVFA concrete. The numerical procedure starts with a hydration model considering both cement hydration and fly ash reaction. The hydration model analyzes cement hydration, the fly ash dilution effect, and the fly ash pozzolanic reaction in cement-fly ash blends. Using the hydration model, the reaction degree of fly ash, CH contents, phases volume fractions, and calcium silicate hydrate contents of hardening cement-fly ash blends are predicted. The compressive strengths of hardening fly ash blended paste are evaluated using calcium silicate hydrate amounts. The efficiency factor of fly ash is dependent on the w/b and is not a constant. The contribution of the fly ash mixtures prepared at a lower w/b was greater than those prepared at a higher w/b. The calculation results from the hydration model are used as input parameters for the carbonation reaction model. By considering the effects of material properties and environmental conditions, the carbonation reaction model analyzes the diffusivity of carbon dioxide and the carbonation depth of HVFA concrete with different curing conditions, different fly ash contents, and different w/b. Increasing the fly ash content or the w/b increases the carbonation depth. Increasing the initial curing periods increases the amounts of carbonatable constituents, decreases the porosity, and decreases the carbonation depth correspondingly. 159

Fig. 9—Carbonation depth of HVFA concrete with different curing periods. (Note: 1 mm = 0.0394 in.) AUTHOR BIOS

Xiao-Yong Wang is an Assistant Professor at Kangwon National University, Chuncheon, South Korea. He received his PhD from Hanyang University, Seoul, South Korea. His research interests include multi-scale modeling of the durability of concrete structures and the anti-seismic behavior of concrete-filled steel tubes. Ki-Bong Park is an Associate Professor at Kangwon National University. He received his PhD from the University of Tokyo, Tokyo, Japan. His research interests include the prediction of thermal and shrinkage cracking in concrete structures using a hydration model and finite element method.

ACKNOWLEDGMENTS

This research was supported by Basic Science Research Program through the National Research Foundation of Korea (NRF), funded by the Ministry of Science, ICT, and Future Planning (No. 2015R1A5A1037548).

REFERENCES

1. Mehta, P. K., and Monteiro, P. J. M., Concrete—Microstructure, Properties, and Materials, McGraw-Hill, New York, 2006, 704 pp. 2. Lam, L.; Wong, Y. L.; and Poon, C. S., “Effect of Fly Ash and Silica Fume on Compressive and Fracture Behaviors of Concrete,” Cement and Concrete Research, V. 28, No. 2, 1998, pp. 271-283. doi: 10.1016/ S0008-8846(97)00269-X

160

3. Papadakis, V. G., “Effect of Supplementary Cementing Materials on Concrete Resistance against Carbonation and Chloride Ingress,” Cement and Concrete Research, V. 30, No. 2, 2000, pp. 291-299. doi: 10.1016/ S0008-8846(99)00249-5 4. Papadakis, V. G.; Fardis, M. N.; and Vayenas, C. G., “Hydration and Carbonation of Pozzolanic Cements,” ACI Materials Journal, V. 89, No. 2, Mar.-Apr. 1992, pp. 119-130. 5. Papadakis, V. G.; Vayenas, C. G.; and Fardis, M. N., “Fundamental Modeling and Experimental Investigation of Concrete Carbonation,” ACI Materials Journal, V. 88, No. 4, July-Aug. 1991, pp. 363-373. 6. Sisomphon, K., and Franke, L., “Carbonation Rates of Concretes Containing High Volume of Pozzolanic Materials,” Cement and Concrete Research, V. 37, No. 12, 2007, pp. 1647-1653. 7. Jiang, L.; Lin, B.; and Cai, Y., “A Model for Predicting Carbonation of High-Volume Fly Ash Concrete,” Cement and Concrete Research, V. 30, No. 5, 2000, pp. 699-702. doi: 10.1016/S0008-8846(00)00227-1 8. Han, S. H.; Kim, J. K.; and Park, Y. D., “Prediction of Compressive Strength of Fly Ash Concrete by New Apparent Activation Energy Function,” Cement and Concrete Research, V. 33, No. 7, 2003, pp. 965-971. doi: 10.1016/S0008-8846(03)00007-3 9. Kim, J. K.; Han, S. H.; and Park, S. K., “Effect of Temperature and Aging on the Mechanical Properties of Concrete: Part II. Prediction Model,” Cement and Concrete Research, V. 32, No. 7, 2002, pp. 1095-1100. doi: 10.1016/S0008-8846(02)00745-7 10. Hwang, K.; Noguchi, T.; and Tomosawa, F., “Prediction Model of Compressive Strength Development of Fly-Ash Concrete,” Cement and

ACI Materials Journal/March-April 2016

Concrete Research, V. 34, No. 12, 2004, pp. 2269-2276. doi: 10.1016/j. cemconres.2004.04.009 11. Park, K. B.; Jee, N. Y.; Yoon, I. S.; and Lee, H. S., “Prediction of Temperature Distribution in High-Strength Concrete Using Hydration Model,” ACI Materials Journal, V. 105, No. 2, Mar.-Apr. 2008, pp. 180-186. 12. Wang, X. Y., “Properties Prediction of Ultra High Performance Concrete Using Blended Cement Hydration Model,” Construction and Building Materials, V. 64, 2014, pp. 1-10. doi: 10.1016/j. conbuildmat.2014.04.084 13. Oh, B. H., and Cha, S. W., “Nonlinear Analysis of Temperature and Moisture Distributions in Early-Age Concrete Structures Based on Degree of Hydration,” ACI Materials Journal, V. 100, No. 5, Sept.-Oct. 2003, pp. 361-370. 14. Matsushita, T.; Hoshino, S.; Maruyama, I.; Noguchi, T.; and Yamada, K., “Effect of Curing Temperature and Water to Cement Ratio on Hydration of Cement Compounds,” Proceedings of 12th International Congress on the Chemistry of Cement, 2007, 12 pp. 15. Maekawa, K.; Chaube, R.; and Kishi, T., Modeling of Concrete Performance: Hydration, Microstructure Formation and Mass Transport, Routledge, London, UK, 1998, 308 pp. 16. Maekawa, K.; Ishida, T.; and Kishi, T., Multi-Scale Modeling of Structural Concrete, Taylor & Francis, London, UK, 2009, 658 pp. 17. Lam, L.; Wong, Y. L.; and Poon, C. S., “Degree of Hydration and Gel/Space Ratio of High-Volume Fly Ash/Cement Systems,” Cement

ACI Materials Journal/March-April 2016

and Concrete Research, V. 30, No. 5, 2000, pp. 747-756. doi: 10.1016/ S0008-8846(00)00213-1 18. Papadakis, V. G., and Tsimas, S., “Supplementary Cementing Materials in Concrete, Part I: Efficiency and Design,” Cement and Concrete Research, V. 32, No. 10, 2002, pp. 1525-1532. doi: 10.1016/ S0008-8846(02)00827-X 19. Papadakis, V. G., and Tsimas, S., “Supplementary Cementing Materials in Concrete, Part II: A Fundamental Estimation of the Efficiency Factor,” Cement and Concrete Research, V. 32, No. 10, 2002, pp. 15331538. doi: 10.1016/S0008-8846(02)00829-3 20. Papadakis, V. G., “Effect of Fly Ash on Portland Cement Systems, Part I: Low-Calcium Fly Ash,” Cement and Concrete Research, V. 29, No. 11, 1999, pp. 1727-1736. doi: 10.1016/S0008-8846(99)00153-2 21. Neville, A. M., Properties of Concrete, John Wiley & Sons, Inc., Hoboken, NJ, 1996, 844 pp. 22. Wang, X. Y.; Lee, H. S.; and Park, K. B., “Simulation of LowCalcium Fly Ash Blended Cement Hydration,” ACI Materials Journal, V. 106, No. 2, Mar.-Apr. 2009, pp. 167-175. 23. Song, H. W.; Kwon, S. J.; Byun, K. J.; and Park, C. K., “Predicting Carbonation in Early-Aged Cracked Concrete,” Cement and Concrete Research, V. 36, No. 5, 2006, pp. 979-989. doi: 10.1016/j. cemconres.2005.12.019

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ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M16

Behavior of Anchored Carbon Fiber-Reinforced Polymer Strips Used for Strengthening Concrete Structures by Wei Sun, James O. Jirsa, and Wassim M. Ghannoum The anchorage of carbon fiber-reinforced polymer (CFRP) strips using CFRP anchors is gaining acceptance in strengthening applications of concrete members. CFRP anchors can fully develop the strength of CFRP strips when adequately detailed. However, parameters that influence the behavior and strength of CFRP strips and anchors are not well understood. In this study, 26 tests on concrete beams were conducted to study the influence of five key parameters on CFRP anchor effectiveness: 1) the width of the anchored CFRP strip; 2) the material ratio of CFRP anchor to CFRP strip; 3) the concrete strength; 4) the length/angle of anchor fan; and 5) the bond condition between a CFRP strip and concrete. Results indicate that narrow anchored CFRP strips developed higher stresses at fracture than wide strips and required smaller anchor material ratios to be fully developed. Test results provide valuable data for designing anchored CFRP strengthening systems. Keywords: anchor(s); bonded; carbon fiber-reinforced polymer; concrete members; strengthening; strip(s); unbonded.

INTRODUCTION Carbon fiber-reinforced polymer (CFRP) materials are widely used to strengthen reinforced concrete structures because they are lightweight, have high strength, and are relatively easy to install. In strengthening applications, CFRP strips are typically attached to the concrete surface using epoxy resin with fibers oriented in the direction in which additional tensile strength is needed. However, if CFRP strips rely exclusively on bond strength with concrete, only about 40 to 50% of the CFRP tensile strength is likely to be developed before debonding occurs.1,2 The tensile strength of CFRP strips in that case are determined by the bond behavior between CFRP and concrete, which are presented using bond stress-slip models of varying complexity in the literature.3-10 The simplest relation used for bond strength and slip is linear,3 while more complex bond-slip models assume bilinear4-6 or even nonlinear7-10 relationships. To prevent CFRP from prematurely debonding from the concrete substrate, anchorage systems have been developed. Mechanically fastened joints involving steel plates and bolts were used to anchor CFRP strips.11 The application of mechanically fastened joints, however, unavoidably introduced practical issues such as stress concentration and corrosion. Recent research has shown that the introduction of CFRP anchors provides an alternate force transfer mechanism so that the strength of the CFRP material can be fully developed after debonding occurs. In a recent study,12 unanchored CFRP U-wraps did not significantly increase the shear strength of reinforced concrete T-beams due to the CFRP strips prematurely debonding from the concrete surface. In the same study, however, shear strength gains exceeding 40% were achieved by anchoring the same CFRP U-wraps using CFRP ACI Materials Journal/March-April 2016

anchors just below the flange. CFRP anchors have also been proven to develop the tensile strength of CFRP strips in flexural strengthening applications,13 as well as to provide continuity in load transfer at locations where CFRP strips cannot run continuously (for example, for columns at the end of a wall14). Many design parameters, the effects of which are not well understood, can affect the behavior and strength of CFRP anchors.15 Inadequately designed CFRP anchors can rupture before the CFRP strips fracture. Many researchers have noted the importance of several anchor details on their efficiency in developing strip strength, mainly: 1) anchor size16,17; 2) details of anchor hole18,19; 3) embedment length, which is the length of the CFRP anchor inserted into concrete20; 4) details of the anchor fan21,22; 5) reinforcing CFRP patch applications23; and 6) anchor layout.17,20 To better assess the effects of anchor details on the performance of CFRP anchors and the strips they develop, 26 tests were conducted on 6 x 6 x 24 in. (152 x 152 x 610 mm) concrete beams strengthened in flexure using anchored CRFP strips. The parameters studied in this research include: 1) the width of CFRP strip; 2) the material ratio of CFRP anchor to CFRP strip; 3) the concrete strength; 4) the length/angle of the anchor fan; and 5) the bond condition between the CFRP strip and concrete. Bond stress/slip relationships between anchored CFRP strips and the concrete substrate were extracted from test results. RESEARCH SIGNIFICANCE The external application of anchored CFRP strips offers an efficient method for repair and strengthening of concrete structures. However, parameters that influence the behavior and strength of CFRP strips and anchors are not well understood. The influence of five key parameters on the strength of anchored CFRP strips was experimentally investigated. Results provide vital information for developing CFRP anchor design and detailing guidelines. EXPERIMENTAL PROGRAM Test specimens The test methodology and specimens used to determine the modulus of rupture of concrete based on ASTM C29324 were adapted for studying the behavior of anchored CFRP strips. Test specimens consisted of concrete beams with ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-072.R2, doi: 10.14359/51688637, received July 15, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

163

Fig. 2—Concrete shear failure of beam without side-face U-wraps.11 Table 1—CFRP material properties

Fig. 1—Beam specimens. (Note: 1 in. = 25.4 mm.) dimension of 6 x 6 x 24 in. (152 x 152 x 610 mm). The small-scale specimens were selected because their size and weight permitted them to be maneuvered easily in the laboratory. As shown in Fig. 1(a), a 1 in. (25.4 mm) deep notch was cut at midspan to ensure flexural cracking occurred at midspan. Holes were drilled 4 in. (102 mm) into the beams for the anchors, and the hole edges were rounded to a radius of 0.5 in. (12.7 mm).25 As shown in Fig. 1(b), a CFRP strip saturated with epoxy resin was then applied to the prepared tension surface of each beam. One square (5 x 5 in. [127 x 127 mm] for 5 in. [127 mm] strips; 3 x 3 in. [76 x 76 mm] for 3 in. [76 mm] strips) CFRP patch was applied at the location of each anchor with fibers oriented perpendicular to the CFRP strip fiber direction. Saturated CFRP anchors were introduced into the holes and fanned over the CFRP patch and strip. An additional square patch was then applied over each anchor with fibers oriented in the direction of the CFRP strip fibers. The patches were added to improve the transfer of force from the strip to the anchor so that strength of the strip was developed before anchor failure.13,23 As shown in Fig. 2, the beam specimens were vulnerable to concrete failure, as they contained no steel reinforcement.19 A shear/flexure crack formed at the location of the drilled hole. Considering the ease of installation of the CFRP material, CFRP strips were used to U-wrap the side faces of the beams to provide additional tensile strength at the section where the anchors were located. U-wraps were discontinuous at midspan and had no influence on the flexural cracking at midspan or on the forces introduced to the anchored CFRP strip. Material properties The same CFRP material was used for CFRP strips and CFRP anchors. Beam specimens, detailed as described 164

Average measured values from three tests

Manufacturer-specified typical test values (using ASTM D3039)

Elastic modulus Ef

15,600 ksi (108 GPa)

15,300 ksi (105 GPa)

Rupture strain

0.0096

0.0093

Rupture stress fCFRP

150 ksi (1034 MPa)

143 ksi (986 MPa)

previously, were used to measure CFRP material properties, except that the strips extended over the ends of the beams and no anchors were used. The properties of this CFRP material are listed in Table 1, in which the average measured CFRP material properties are compared with manufacturerspecified values. The manufacturer-specified values obtained from direct tensile tests in accordance with ASTM D3039 were nearly identical to measured values and will be used to determine stresses from measured strains in this study.25 Variables The width of CFRP strips, anchor fan length, and anchor fan angle are illustrated in Fig. 3. As shown in Fig. 1, CFRP anchors were introduced into the holes and fanned over the CFRP strip. The anchor fan length refers to the radius of the fan, and anchor fan angle refers to the angle of fan, as shown in Fig. 3. The anchor material ratio (AMR), which is the material ratio of CFRP anchor to CFRP strip at any given section, is illustrated in Fig. 3. Because anchor and strip were made by the same CFRP material, the value of AMR in this study is equal to wanchor tf anchor /wstrip tf strip, in which wanchor and tf anchor are the width and thickness of anchor, respectively; similarly, wstrip and tf strip are the width and thickness of strip at any given section, respectively. To investigate the load transfer mechanism from strip to anchors with and without strip bonding to the concrete substrate, a plastic film was placed between the concrete surface and the CFRP strip to prevent bond in some tests. The range of geometric and material properties of the beam specimens tested are as follows: • Concrete strength at time of failure fc′: 5.4 or 11.5 ksi (37 or 79 MPa) • Strip width (SW): 3 or 5 in. (76 or 127 mm) • Anchor material ratio (AMR) is ratio of anchor to strip material = 1.06 to 2.0 ACI Materials Journal/March-April 2016

Fig. 3—Specimen details. (Note: 1 in. = 25.4 mm.)

Fig. 4—Test setup. (Note: 1 in. = 25.4 mm.) • • •

Anchor fan angle (FA): 37 to 64 degrees Anchor fan length (FL): 2.4 to 7.5 in. (61 to 191 mm) Bonded application (BA) and unbonded application (UA) Additional details about the experimental program can be found in Table 2 and Reference 25. Test setup To develop tensile force on the anchored CFRP strip, the concrete beam was loaded at midspan through a spherical head and supported by rockers and threaded rods, as shown in Fig. 4. The threaded rods transferred the force from the test beam to a reaction beam. The load cell, hydraulic ram, and spherical head were placed between the reaction beam and the test specimen. Two linear voltage displacement transducers (LVDTs) were used to record the displacements at midspan and at a support to determine midspan beam deflection. The threaded rods were flexible and prevented the development of axial forces in the beams as deformations increased. Recently, digital image correlation (DIC) systems have been introduced in structural engineering to measure surface deformations.26-31 A high-resolution DIC system developed by Sokoli et al.32 was used in this study to record the threedimensional (3-D) movements of targets affixed to the tension surface of the beam specimens. A typical test setup of the DIC system is shown in Fig. 5. This setup was placed on a table at a height that matched that of the cameras used ACI Materials Journal/March-April 2016

in the DIC system. The beam was loaded horizontally so that the cameras faced the tension surface. Beam deflection was calculated as the relative displacement in the z-direction (perpendicular to beam surface) between the targets at midspan and those at the ends and compared with LVDT readings (Fig. 6(a)). Good correlation between deformations recorded by LVDTs and the DIC system were observed (Fig. 6(a)). Small discrepancies can be attributed to the slight difference in the locations where measurements were taken. The x-component strain εx in a given frame number (i) is calculated as the change in x-direction (longitudinal direction of beam surface) distance (Δlxi) between two targets divided by the original x-direction distance (Δl) between those two targets

εx =

∆lxi (1) ∆l

Excellent agreement was observed between strain measurements recorded using strain gauges placed 1 in. (25.4 mm) from strip edge and the DIC measurements from nearby targets using Eq. (1) (Fig. 6(b)). (It is important to note that strains do not match exactly between DIC and strain gauge readings because the locations monitored were slightly different. DIC strain measurements were calculated from targets around the strain gauges). 165

Surface targets can be used as nodes of quadrilateral planar elements. The in-plane strains of the elements can be calculated from the coordinate changes of four targets Table 2—Test details AMR

fc′, ksi (MPa)

FA, degrees

FL, in.

B5H2Ma

2.0

11.5 (79)

45

6

2

B5H2Mb

2.0

11.5 (79)

45

6

3

B5H1.4Ma

1.41

11.5 (79)

45

6

4

B5H1.4Mb

1.41

11.5 (79)

45

6

5

B5H1.4Md

1.41

11.5 (79)

45

6

6

B5H1.4Sb

1.41

11.5 (79)

64

4

7

B5H1.4La

1.41

11.5 (79)

37

7.5

8

B5H1.4Lb

1.41

11.5 (79)

37

7.5

9

B5L1.4Ma

1.41

146.7

45

6

10

B5L1.4Mb

1.41

134.7

45

6

11

B5L1.4Mc

1.41

157.7

45

6

12

B5H1Ma

1.06

137.5

45

6

13

B5H1Mb

1.06

144.8

45

6

14

B5H1Mc

1.06

147.6

45

6

15

B5L1Ma

1.06

142.1

45

6

16

B5L1Mb

1.06

104.5

45

6

17

B5L1Mc

1.06

125.6

45

6

18

B5L1Md

1.06

135.7

45

6

19

B5L1Me

1.06

141.2

45

6

20

B5L1Mg

1.06

102.7

45

6

21

B3H1.4Sa

1.41

11.5 (79)

64

2.4

22

B3H1.4Sb

1.41

11.5 (79)

64

2.4

23

B3H1.4Ma

1.41

11.5 (79)

45

3.6

24

B3H1.4Mb

1.41

11.5 (79)

45

3.6

No.

Specimen

1

SW, in.

5

3

25

B3H1.4La

1.41

11.5 (79)

37

4.5

26

B3H1.4Lb

1.41

11.5 (79)

37

4.5

Notes: Specimen nomenclature: First character B or U refers to bonded (B) or unbonded (U) specimens; second number refers to 5 in. (5) or 3 in. (3) wide CFRP strip; third character refers to concrete strength as H (higher, 11.5 ksi) and L (lower, 5.4 ksi); fourth number refers to anchor material ratio as 2 (2.0), 1.4 (1.41), and 1 (1.06); fifth character refers to anchor fan angle as S (64 degrees), M (45 degrees), and L (37 degrees); and last character refers to unique test ID; 1 in. = 25.4 mm.

assuming linear strain profiles and used to produce CFRP strip surface-strain contours. In Fig. 7, the contours of the x-direction (longitudinal) strains are plotted for two tests at various loading stages. The strain contours allowed the visualization of surface-strain distributions and concentrations. In Fig. 7, the locations of anchor fans and rectangular patches are highlighted. Typical test behavior Two major failure modes were observed for anchored specimens: CFRP strip fracture when the tensile strength of CFRP strip is realized (Fig. 8); and anchor rupture due to an insufficient amount of CFRP material in anchors, leading to rupture of anchors before fracture of strips (Fig. 8). Load-deflection responses for two test specimens are presented in Fig. 8. The highlighted test specimens had identical parameters except for the AMR. The typical loaddeflection responses of both strengthened beams had an almost linear relation prior to flexural cracking (Fig. 8), which suggested that CFRP strips remain fully bonded to the concrete substrate until cracking. Following cracking, the load-deflection curves experienced a gradual softening due to strip debonding (Fig. 8) until most of the load was transferred to the anchors and a nearly linear load-deflection response was again observed (Fig. 8). If the CFRP anchors were properly designed and installed, fracture of the CFRP strip was observed at ultimate load. Otherwise, anchor rupture at the anchor-hole edge was observed at a lower ultimate load than required to fully develop the tensile strength of the CFRP strip. In all cases, failure was brittle and sudden. In tests with bonded strip, uniformly distributed strains were typically observed prior to beam cracking or 25% of the ultimate load (Fig. 7). After flexural cracking, debonding between the CFRP strip and the concrete substrate initiated at midspan and propagated toward the CFRP anchors with increasing applied load, as can be deduced from increasing CFRP strains spreading away from midspan with increasing load (Fig. 7). Debonding mechanism Bond stress-versus-slip relationships can be used in computational models to numerically study the load-transfer mechanism from CFRP strips to the concrete substrate and

Fig. 5—Setup of DIC system cameras to monitor specimen tension face deformations. 166

ACI Materials Journal/March-April 2016

Fig. 6—Comparisons between: (a) midspan deflections measured by DIC system and LVDTs on Specimen B5H1.4Sb; and (b) strains measured by DIC system and strain gauges on Specimen B5L1Mc. (Notes: 1 in. = 25.4 mm; 1 kip = 4.45 kN.)

Fig. 7—Progression of longitudinal strain contours for two typical tests (left: B5H1.4Ma; and right: B5H1Mc). anchors. The relation between bond stress and slip was extracted from the test data. The change in tensile force along a CFRP strip is generated through bond forces at the interface between the strip and the concrete substrate (Fig. 9). The bond stress developed between a CFRP strip and concrete can be determined from strain measurements as follows. The change in strip tensile force between targets i and i + 3 in a given row of targets can be written as

Ef is the manufacturer-specified modulus of elasticity of the CFRP strip (Table 1); tf is the specified thickness of the CFRP strip; bt is the center-to-center distance between target rows = 0.5 in. (12.7 mm); and ΔFi,i+3 is the change in tensile force in the CFRP strip over distance ΔXi,i+3 within width bt. Thus, solving for τi,i+3

∆Fi,i+3 = τi,i+3∆Xi,i+3bt (2)

The slip between a CFRP strip and the concrete substrate is evaluated as the cumulative x-direction elongation between the locations of targets at the edge of the anchorage region where no slip occurs, and the target locations where slip is evaluated



The change in strip tensile force can also be written as

∆Fi,i+3 = (εi+2,i+3 – εi,i+1)Ef  tf  bt (3)

where ∆Xi,i+3 is the distance between two adjacent pairs of targets starting at target i and ending at target I + 3; τi,i+3 is the average bond stress over the shaded area bounded by targets i and i + 3 and the midspan to the adjacent rows of targets; εi,i+1 is the strain measured between the first two targets; εi+2,i+3 is the strain measured between the last two targets; ACI Materials Journal/March-April 2016





τ i ,i + 3 =

ε i + 2 , i + 3 − ε i , i +1

slipi ,i + 3 =

∆X i ,i + 3

E f t f (4)

∆i + ∆i+3 − ∆ 0 (5) 2

where Δi is the x-direction displacement of i-th target; Δi+3 is the x-direction displacement of i+3-th target; and Δ0 is the x-direction displacement of the target at location of zero slip in a target row. 167

Fig. 8—Load-deflection plots for typical strip fracture (B5H1.4Ma) and anchor rupture tests (B5L1Mc). (Note: 1 in. = 25.4 mm; 1 kip = 4.45 kN; 1 ksi = 6.89 MPa.)

Fig. 9—Tensile force transfer from CFRP strip to concrete substrate and typical extracted bond stress-versus-slip relationship of Specimen B5H1.4Sb. (Note: 1 in. = 25.4 mm; 1 ksi = 6.89 MPa.) Bond stress and slip values evaluated at targets across a strip width from the midspan extending four targets toward anchor fan edges were averaged. A typical resulting average bond stress-versus-slip relation is shown in Fig. 9. In that relation, bond stress increased up to 0.51 ksi (3.5 MPa) at a slip of 0.003 in. (0.08 mm). After that, bond stress decreased to zero when slip reached 0.01 in. (0.25 mm). Bond-versus-slip relations were extracted for tests with 5 in. (127 mm) and 3 in. (76 mm) strips separately, and simplified bond-versus-slip relations between CFRP strips and concrete were produced (Fig. 10). The simplified bondversus-slip relations are linear up to peak bond stress. Then, 168

a linear degrading behavior represents the response from the peak to the slip at which the CFRP strip is completely debonded from the concrete substrate and bond stress is zero. The peak bond stress and its corresponding slip, as well as the slip at zero stress shown in Fig. 10, represent average values from pertinent tests. As can be seen in the Fig. 10, higher-strength concrete generates a higher peak bond stress but lower slip at peak stress than the lowerstrength concrete. The higher peak bond stress and lower slip at peak stress make the ascending slope of the bondversus-slip relation stiffer for higher-strength concrete. For the degrading branch, a steeper slope was also observed for ACI Materials Journal/March-April 2016

Fig. 10—Concrete-CFRP bond stress-versus-slip relationships extracted from DIC strain data. (Note: 1 in. = 25.4 mm, 1 ksi = 6.89 MPa.) Table 3—Experimental results for effect of concrete strength on strip strength Common parameters fc′, ksi (MPa) σufx, ksi (MPa) BA SW = 5 in. (127 mm) AMR = 1.41 FA = 45 degrees Failure mode = strip fracture

Average σufx, ksi (MPa)

141 (972) 11.5 (79)

143 (986)

142 (979)

143 (986) 5.4 (37)

145 (1000) 134 (924)

140 (965)

specimens with high-strength concrete compared with that of specimens with normal-strength concrete. It is noteworthy that the peak bond stresses extracted from test data match concrete tensile strengths estimated using ACI 318-14,33 which range from ft = 6 to 7.5√fc′ (in psi units) (ft = 0.5 to 0.62√fc′ [in MPa units]). For instance, the peak bond stress between 5 in. (127 mm) strips and 5.4 ksi (37 MPa) concrete of 0.45 ksi (3.1 MPa) corresponds to 6.1√fc′ (in psi units) (0.51√fc′ [in MPa units]). Similarly, the peak bond stresses for other cases ranged from 8.3 to 8.7√fc′ (in psi units) (0.69 to 0.72√fc′ [in MPa units]). Test results therefore indicate that the peak bond stress between anchored CFRP strips and concrete may reasonably be estimated as the concrete tensile strength for general anchored CFRP strip applications.34 EXPERIMENTAL RESULTS Based on 26 beam tests, the effects of the following parameters—concrete strength, anchor fan length/angle, width of CFRP strips, ratio of CFRP anchor material to CFRP strip material, and bond between CFRP and concrete on strip and anchor strengths—are discussed in this section. Effects of concrete strength Five tests that failed by strip fracture were evaluated to study the impact of concrete strength on CFRP strip strength. In Table 3, the strip stress at midspan σufx was used to evaluate the effect of concrete strength. The strip stress at midspan σufx is evaluated at ultimate load and equal to Ff mid /ACFRP. ACFRP is the cross-sectional area of the CFRP strip, and Ff mid is the strip force at midspan, which is calcuACI Materials Journal/March-April 2016

Fig. 11—Beam equilibrium. (Note: 1 in. = 25.4 mm.) lated by equilibrium at ultimate load Pult using the ACI 318-1433 stress block approach when the depth of nonlinear compression zone on 5.4 and 11.5 ksi (783 and 1668 psi) concrete are 0.67 and 0.22 in. (17 and 6 mm), respectively. Beam forces are illustrated in Fig. 11. As shown in Table 3, the concrete strength did not have a significant effect on σufx. Seven specimens that failed by anchor rupture were evaluated to study the impact of concrete strength on anchor strength. As shown in Table 4, σufx averaged over specimens with the same concrete strength was higher for specimens with high-strength concrete than specimens with lowerstrength concrete. The high-strength concrete resulted in an increase of approximately 10% in the ultimate strip stress at anchor failure. Effects of anchor fan length/anchor fan angle Six tests that failed by strip fracture were evaluated to study the impact of anchor fan length/angle on strip strength. Only the anchor fan length/angle varied in this group. To effectively develop the strength of CFRP strips, CFRP anchors were fanned out across the width of CFRP strips. Because strip width was kept the same in each group, the length of the anchor fan determines the anchor fan angle. As shown in Table 5, all strips fractured at an ultimate strip stress σufx larger than the expected tensile strength provided by the manufacturer (143 ksi [986 MPa]). Overall, increasing the fan angle from 37 to 64 degrees did not produce a significant change in the ultimate strip stress at strip fracture. (No conclusion could be made concerning the effects of anchor fan geometry on anchor strength due to insufficient data from tests sustaining anchor failures and having a range of fan geometries.) 169

Table 4—Experimental results for effect of concrete strength on anchor strength Common parameters

fc′, ksi (MPa)

σufx, ksi (MPa)

Average σufx, ksi (MPa)

134 (924) BA SW = 5 in. (127 mm) AMR = 1.06 FA = 45 degrees Failure mode = anchor rupture

11.5 (79)

141 (972)

140 (965)

144 (993) 142 (979) 5.4 (37)

104 (717) 125 (862)

127 (876)

Fig. 12—Selected area and targets used to measure strip strains.

135 (931)

Table 5—Experimental results for effect of fan geometry on strip strength Common parameters Fan length/angle 2.4 in. (61 mm)/ BA 64 degrees SW = 3 in. (76 mm) fc′ = 11.5 ksi 3.6 in. (91 mm)/ (79 MPa) 45 degrees AMR = 1.41 Failure mode = strip 4.5 in. (114 mm)/ fracture 37 degrees

σufx, ksi (MPa) 154 (1062) 174 (1200) 183 (1262) 154 (1062) 186 (1282) 148 (1020)

Average σufx, ksi (MPa) 164 (1131) 169 (1165) 167 (1151)

Table 6—Experimental results for effect of strip width on strip strength Common parameters BA fc′ = 11.5 ksi (79 MPa) AMR = 1.41 FA = 45 degrees Failure mode = strip fracture

Strip width, in. (mm)

σufx, ksi (MPa)

Average σufx, ksi (MPa)

141 (972) 5 (127)

143 (986)

142 (979)

143 (986) 3 (76)

183 (1262) 154 (1062)

169 (1165)

Effects of width of CFRP strip Five tests were compared in Table 6 to investigate the impact of strip width on strip strength. All tests failed due to strip fracture. In tests with 5 in. (127 mm) wide CFRP strips, the ultimate strip stress σufx at fracture was nearly the same and significantly lower than the value of σufx measured from tests with 3 in. (76 mm) wide strips. The average ultimate strip stress σufx was larger in the narrower strips than the wider ones. All tests reached or exceeded the expected CFRP strip stress at failure (143 ksi [986 MPa]). The surface longitudinal strains between two adjacent targets at 98% of specimen ultimate load εusx were measured over the CFRP strip area bounded by anchor fan ends, as illustrated in Fig. 12. Mean and maximum values of εusx over the area considered are presented in Fig. 13. As shown in Fig. 13, the maximum longitudinal strip strains just prior to strip fracture were higher for 5 in. (127 mm) strips, and the differences between the maximum and mean strip strains at 98% of the ultimate load were also greater. Thus, the wider strips were observed to experience both higher localized maximum strip strains and higher differences between 170

Fig. 13—Comparison of mean and maximum strains at 98% ultimate load (εusx) for different strip widths (BA; SW = 5 in. [127 mm]; fc′ = 11.5 ksi [79 MPa]; AMR = 1.41.) maximum and mean strip strains. These findings indicate that with a wider CFRP strip, strain distributions across the strip area were less uniform and exhibited higher peak strains. Because CFRP is a brittle material, higher local strains in wider strips may be the cause of their observed lower strength compared with the 3 in. (76 mm) narrower strips. Effects of material ratio of CFRP anchor to CFRP strip Anchors with AMRs of 1.06, 1.41, and 2.0 were studied to determine the effects of AMR on strip and anchor strengths. Fourteen tests were conducted on bonded 5 in. (127 mm) strips with AMRs of 1.06, 1.41, and 2.0. Another 10 tests were conducted on beams with bonded 3 in. (76 mm) strips using anchors with material ratios of 1.06 or 1.41. Figure 14 highlights specimen failure modes and the ratio of ultimate loads applied at specimen failure to the expected specimen strength assuming a strip fracture mode and manufacturer specified CFRP material properties. As shown in Fig. 14, with an AMR of 1.41, some 5 in. (127 mm) wide CFRP strips did not reach expected strength. For tests with 3 in. (76 mm) strips, shown in Fig. 14, all 10 tests exceeded the expected strength of the CFRP strips regardless of the AMR. Because the desired mode of failure is strip fracture, it is reasonable to suggest that an AMR of 1.4 should be used to reach fracture of 3 in. (76 mm) strips and an AMR of 2.0 should be used for 5 in. (127 mm) strips. In Fig. 15, mean and maximum values of strains between targets in the area shown in Fig. 12 are indicated at 95% of the expected load at failure εesx. As can be seen in Fig. 15, at the same applied load, anchors with a material ratio of 2.0 had significantly reduced maximum strip strains and smaller differences between maximum and mean strip strains, compared with anchors having an AMR of 1.41. Therefore, anchors with a larger cross section are observed to achieve, ACI Materials Journal/March-April 2016

Fig. 14—Failure modes for tests with 5 and 3 in. (127 and 76 mm) wide strips. at a given load, more even strain distributions and lower maximum strains than smaller anchors. Such favorable strain distributions resulted in an increase in the ultimate strip stress at fracture when larger anchors were used. Effects of bonded versus unbonded applications A bonded test designates that epoxy resin was used as the interfacial material to bond the CFRP strip to the concrete substrate. Unbonded tests indicate that a plastic film was placed between the CFRP strip and the concrete substrate to simulate the behavior of a completely debonded strip. Four directly comparable tests were conducted with the bonding of the CFRP strip to the concrete using epoxy or using a plastic film. In all tests, the AMR was 1.41, strips were 5 in. (127 mm) wide, anchor fans were 6 in. (152 mm) long, and high-strength concrete was used. As shown in Fig. 16, unbonded specimens failed at ultimate loads lower than the expected load at failure (which was 16 kip [71 kN]) and the anchors ruptured. In bonded applications, the bond between the CFRP strips and concrete seems to have increased the apparent strength at anchor fracture. The CFRP-concrete bond may distribute anchor stresses more evenly at the anchor area. SUMMARY AND CONCLUSIONS In this study, failure modes and ultimate load and strain measurements were used to evaluate the effects of five parameters on the performance of anchored CFRP strips. The main findings from the study with respect to those five parameters are listed as follows. 1. Test results indicate that to fracture a 5 in. (127 mm) wide CFRP strip (strip fracture), the AMR should be no less than 2.0. Increasing the AMR from 1.41 to 2.0 reduced strain concentrations, resulting in higher average ultimate strip stresses at fracture. An AMR not less than 1.41 is recommended for 3 in. (76 mm) strips. 2. Test results show that increasing the width of CFRP strips resulted in higher local peak strains and lowered the average stress at fracture of the strip. 3. Increasing concrete strength increased the bond strength between CFRP strips and the concrete substrate. Thus, debonding of the CFRP strip occurred at a higher load for higher-strength concrete. A higher concrete strength was found to slightly increase the strength of CFRP anchors ACI Materials Journal/March-April 2016

Fig. 15—Comparison of mean and maximum strains at 95% expected load at failure (εesx) for different anchor-material ratio (BA; SW = 5 in. [127 mm]; fc′ = 11.5 ksi [79 MPa]; FA = 45 degrees).

Fig. 16—Typical load-versus-deflection responses for tests with different bond condition and AMR = 1.41. (Notes: 1 in. = 25.4 mm; 1 kip = 4.45 kN.) embedded in it. The peak bond stress between anchored CFRP strips and the concrete may reasonably be estimated as the concrete tensile strength. 4. To fully develop the tensile strength of a CFRP strip, an anchor fan angle less than 64 degrees is recommended for anchor design. The application of an anchor fan angle smaller than 64 degrees (down to 37 degrees), however, had no significant effect on the strength and behavior of the CFRP strengthening system. 171

5. Adequately bonding the CFRP strips to the concrete substrate helped to transfer tensile forces from CFRP strips to CFRP anchors, and prevented premature anchor rupture due to strain concentrations. AUTHOR BIOS

Wei Sun is an Associate Professor in the School of Civil Engineering and Mechanics at Lanzhou University, Lanzhou, Gansu, China. He received his BS and MS from Shenyang Jianzhu University, Shenyang, China, and his PhD from the University of Texas at Austin, Austin, TX. His research interests include strengthening structures with carbon fiber-reinforced polymer materials. ACI Honorary Member James O. Jirsa is the Janet S. Cockrell Centennial Chair in Engineering in the Department of Civil, Environmental, and Architectural Engineering at the University of Texas at Austin. He is a Past President of ACI and a member of ACI Committee 318, Structural Concrete Building Code. ACI member Wassim M. Ghannoum is an Assistant Professor in the Department of Civil, Environmental, and Architectural Engineering at the University of Texas at Austin. He is Chair of ACI Committee 369, Seismic Repair and Rehabilitation, and a member of ACI Subcommittees 318-R, High-Strength Reinforcement (Structural Concrete Building Code), and 440-F, FRP-Repair-Strengthening. His research interests include extending the life span and increasing the resilience to damage of concrete structures through the application of novel structural materials and retrofit techniques.

ACKNOWLEDGMENTS

The support of the Texas Department of Transportation for Projects 0-6306 and 0-6783 is gratefully acknowledged. The contents of this paper reflect the views of the authors, who are responsible for the facts and the accuracy of the data presented herein. The contents do not necessarily reflect the official view or policies of the Federal Highway Administration or the Texas Department of Transportation (TxDOT). This paper does not constitute a standard, specification, or regulation. The authors express their thanks to the students, faculty, and staff at Ferguson Structural Engineering Laboratory for their assistance with the project.

REFERENCES

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12. Kim, Y.; Quinn, K. T.; Ghannoum, W. M.; and Jirsa, J. O., “Strengthening of Reinforced Concrete T-Beams Using Anchored CFRP Materials,” ACI Structural Journal, V. 111, No. 5, Sept.-Oct. 2014, pp. 1027-1036. doi: 10.14359/51686805 13. Kim, I., “Use of CFRP to Provide Continuity in Existing Reinforced Concrete Members Subjected to Extreme Loads,” PhD dissertation, Department of Civil, Environmental and Architectural Engineering, University of Texas at Austin, Austin, TX, 2008, 478 pp. 14. Kobayashi, K.; Fuji, S.; Yabe, Y.; Tsukagoshi, H.; and Sugiyama, T., “Advanced Wrapping System with CF Anchor—Stress Transfer Mechanism of CF Anchor,” 5th International Symposium on Fiber-Reinforced Polymer (FRP) Reinforcement for Concrete Structures, Cambridge, UK, 2001, pp. 379-388. 15. Smith, S. T., “FRP Anchors: Recent Advances in Research and Understanding,” Asia-Pacific Conference on FRP in Structures, 2009, pp. 35-44. 16. Niemitz, C.; James, R.; Breña, S.; “Experimental Behavior of Carbon Fiber-Reinforced Polymer (CFRP) Sheets Attached to Concrete Surfaces Using CFRP Anchors,” Journal of Composites for Construction, ASCE, V. 14, No. 2, 2010, pp. 185-194. 17. Smith, S. T.; Zhang, H.; and Wang, Z., “Influence of FRP Anchors on the Strength and Ductility of FRP-Strengthened RC Slabs,” Construction and Building Materials, V. 49, 2013, pp. 998-1012. doi: 10.1016/j. conbuildmat.2013.02.006 18. Pham, L. T., “Development of a Quality Control Test for Carbon Fiber Reinforced Polymer Anchors,” master’s thesis, University of Texas at Austin, Austin, TX, 2009, 87 pp. 19. Huaco, G., “Quality Control Test for Carbon Fiber Reinforced Polymer (CFRP) Anchors for Rehabilitation.” master’s thesis, University of Texas at Austin, Austin, TX, 2009, 597 pp. 20. Orton, S. L., “Development of a CFRP System to Provide Continuity in Existing Reinforced Concrete Structures Vulnerable to Progressive Collapse,” PhD dissertation, Department of Civil, Environmental and Architectural Engineering, University of Texas at Austin, Austin, TX, 2007. 21. Breña, S.F., and McGuirk, G. N., “Advances on the Behavior Characterization of FRP-Anchored Carbon Fiber-Reinforced Polymer (CFRP) Sheets Used to Strengthen Concrete Elements,” International Journal of Concrete Structures and Materials, V. 7, No. 1, 2013, pp. 3-16. 22. Zhang, H. W., and Smith, S. T., “Influence of FRP Anchor Fan Configuration and Dowel Angle on Anchoring FRP Plates,” Composites Part B: Engineering, V. 43, No. 8, 2012, pp. 3516-3527. doi: 10.1016/j. compositesb.2011.11.072 23. Kalfat, R.; Al-Mahaidi, R.; and Smith, S. T., “Anchorage Devices Used to Improve the Performance of Reinforced Concrete Beams Retrofitted with FRP Composites State-of-the-Art Review,” Composites for Construction, V. 17, No. 1, 2013, pp. 14-33. 24. ASTM C293-07, “Standard Test Method for Flexural Strength of Concrete Using Simple Beam With Center-Point Loading,” ASTM International, West Conshohocken, PA, 2007, 3 pp. 25. Sun, W., “Behavior of Carbon Fiber Reinforced Polymer (CFRP) Anchors Strengthening Reinforced Concrete Structures,” PhD dissertation, University of Texas at Austin, Austin, TX, 2014, 250 pp. 26. Choi, H. S.; Cheung, J. H.; Kim, S. H.; and Ahn, J. H., “Structural Dynamic Displacement Vision System Using Digital Image Processing,” NDT&E International, V. 44, No. 7, 2011, pp. 597-608. doi: 10.1016/j. ndteint.2011.06.003 27. Helfrick, M. 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ACI Materials Journal/March-April 2016

ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M17

Effect of Dosage of Fly Ash and NaOH on Properties of Pisha Sandstone-Based Mortar by Changming Li, Tingting Zhang, and Lijiu Wang This paper investigates the effectiveness of NaOH concentration and fly ash dosage on the mechanical properties, pore structure, water resistance, and water permeability of a new mortar manufactured by using a special kind of sandstone and fly ash. Mechanical properties were evaluated by means of compressive strength. Mercury intrusion porosimetry (MIP) was employed to investigate the pore structure and pore size distribution. A water permeability test was carried out to find out the permeation characteristics of Pisha sandstone-based mortar. Thermogravity and differential scanning calorimetry (TG-DSC), Fourier transform infrared (FTIR) analysis, and scanning electron microscopy (SEM) were used to analyze the reaction products, indentify the phases of the reaction products, and observe the morphology, respectively. Test results revealed that NaOH concentration and fly ash dosage had significant effects on the mechanical properties, pore structure, and water permeability of Pisha sandstone-based mortar. The specimen (NaOH concentration 3.1 mol/L, fly ash wt.15%) exhibited the highest compressive strength (19.32 MPa [2801 psi]), water resistence (0.76) and lowest water permeability value (5.1 × 10–8 m/s [16.73 × 10–8 ft/s]) at 90 days. Keywords: compressive strength; microstructure; Pisha sandstone-based mortar; pore structure; water permeability.

INTRODUCTION Pisha sandstone (PS) is a special kind of sandstone formed during the Tertiary period; it is hard when it is dry but would collapse when immersed in water because its cementitious material—carbonate mineral calcite—is easy to dissolve in water.1,2 Due to PS’s unsatisfactory bonded mechanism, soil erosion occurs frequently during and after rainstorms; the soil erosion rate of PS area is very high (over 20,000  t∙km–2·yr–1 [50,995 lt·sq.mi–2·yr–1]).3 Although the area of PS only accounts for 2.6% of the total Loess Plateau area of China, its coarse sediment yield accounts for 30% of the total coarse sediment into the upper-middle reaches of the Yellow River.4,5 Soil and water loss has become the main environmental disaster to the local agricultural production and environment, and a great quantity of sediment was deposited on the Yellow River course and raised the altitude of the riverbed. PS became the main source of sediment into the Yellow River.6,7 Check dams were a very important engineering measure for soil and water conservation. There is considerable research that has been carried out to understand the check dam’s effect of trapping sediment, and the results showed that the check dam achieved the highest efficiency in trapping sediment among the all ecological and engineering measures.8,9 However, many of the existing check dams and sea buckthorn flexible dams are often damaged and broken down by flood due to the low strength and defect of dam material (the dam material, PS, would collapse in water).10 ACI Materials Journal/March-April 2016

Fly ash is a by-product of coal-burning power plants and is usually considered a waste material. While more than 600  million tons of fly ash is generated each year worldwide, 80% is disposed of in landfills.11 With pozzolanic and cementitious properties, it has been used as a substitute for cement in concrete and mortar.12 PS contains sufficient amounts of reactive alumina and silica, and its main mineral composition includes quartz, feldspar, and montmorillonite; the feasibility of producing structural material by using PS via the alkali activation process had been studied by Li et al.,2 and their conclusions showed that it is feasible to synthesize geopolymers by using PS. The aim of this study is to investigate the relationship between mixture proportions, pore structure, water resistence, water permeability, and microstructure of Pisha sandstone-based mortar (PSM). The studied mixture parameters were: NaOH concentration, fly ash dosage, and curing age. The influences of NaOH concentration and fly ash dosage on the compressive strength, pore structure, water permeability, and microstructure of PSM were investigated to evaluate the properties of the materials. Hoping that a new mortar or concrete material would be produced by the suggested method, the exploration could offer suggestions for exploitation of PSM in engineering of check dams and building bricks in the future. RESEARCH SIGNIFICANCE PS is a special kind of sandstone; it was the main source of sediment into the river. The by-product, fly ash, is usually considered a waste material, and over 600 million tons of fly ash is generated each year worldwide, causing environmental problems. Thus, it is a beneficial exploration to manufacture structural materials using PS and fly ash by means of alkali activation. In this study, the relationships among the compressive strength, pore structure, water resistance, water permeability, and microstructure of PSM were studied, and the engineering characteristics of PSM was investigated for further evaluation. The results are expected to promote the exploration and use of PS and fly ash, and the application of PSM in engineering in the future. EXPERIMENTAL PROCEDURE Materials The PS used in this study was collected from city of Ordos in the northern Loess Plateau of China. The chemACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-099, doi: 10.14359/51688462, received April 1, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

173

ical compositions of raw materials are presented in Table 1. The raw materials used to synthesize PSM include PS, sodium silicate solution consisting of 14 wt.% NaOH and 27 wt.% SiO2, and sodium hydroxide (99% purity quotient), while those used to synthesize PS and fly ash-based mortar (PSFM) are PS, Class F fly ash, and sodium silicate solution consisting of 14 wt.% NaOH and 42 wt.% SiO2 and sodium hydroxide (99% purity quotient). The PS was air-dried, homogenized, and pulverized until all solids passed a No. 18 mesh (1000 µm opening) sieve to facilitate geopolymerization reaction and minimize the influence of compositional variation on the synthesis. Table 1 summarizes the chemical composition of the PS and fly ash, while Fig. 1 shows the feature and detailed size distributions of PS. Sample preparation Table 2 summarizes the materials mixture used to prepare the mortar. For all the samples, the ratio of activator solution Table 1—Chemical composition of raw materials by X-ray fluorescence Items

Pisha sandstone

Fly ash

SiO2,%

65.64

57.45

Al2O3,%

14.35

27.03

CaO,%

8.02

3.11

Na2O,%

1.07

0.43

K2O,%

1.81

2.60

MgO,%

3.95

0.97

Fe2O3,%

4.03

5.60

P2O5,%

0.10

0.27

LOI

1.03

2.54

Note: LOI is loss on ignition at 1000°C (1832°F); fly ash satisfied JIS R5213.

to solid mixture was kept at a constant value of 0.15, and the total mass of the samples was kept at a constant value of 392.5 g (13.85 oz). For the synthesis of PSM and PSFM, sodium hydroxide (NaOH) was first dissolved in deionized water for 5 minutes to make an NaOH solution with a concentration of 1.35 to 3.1 mol/L. Then, the as-received sodium silicate solution was added to the NaOH solution, followed by mixing for 5 minutes. The mixture of PS and fly ash was added to this solution, followed by mixing for 15 minutes with a magnetic stir bar to achieve complete mixing between the solid and solution, resulting in the formation of mortar precursor. To make regularly shaped specimens for mechanical testing, the mortar precursor were poured into cylindrical steel molds with an inner diameter of 5 cm (1.968 in.) and height of 15 cm (5.904 in.), and the mixture was pressed to specimens with a diameter of 5 cm (1.968 in.) and height of 10 cm (3.936 in.) (that is, an aspect ratio of 2.0 to minimize the end effects) on a hydraulic testing machine. To ensure repeatability, three specimens were prepared for each type of mortar of a predesigned composition (for example, NaOH concentration, fly ash content) at each curing duration. Figure 2 shows the feature of PSM specimens. All samples were sealed using a layer of plastic bag and were kept in a climate room with a temperature of 25 ± 2°C (77 ± 3.6°F) for 3, 7, 28, and 90 days. Characterization techniques Compressive strength—Unconfined compression strength tests were performed on cured cylindrical specimens using an electronic universal testing machine with a 100 kN (22.48 kip) capacity and a constant displacement rate of 0.5%/min (ASTM C39/C39M).13 During testing, a very thin layer of lubricant coating was applied to the two ends of each specimen to minimize the friction and hence shear

Table 2—Mixture proportions of PSM and PSFM Mixture

*

Activator solution

Sample

Pisha, wt.%

Fly ash, wt.%

NaOH concentration

Na2SiO3, wt.%

s/b*, wt.%

PS PSM1.35 PSM2.3 PSM3.1 PSFM5 PSFM10 PSFM15

87 87 87 87 82 77 72

0 0 0 0 5 10 15

— 1.35 mol∙L–1 2.3 mol∙L–1 3.1 mol∙L–1 2.3 mol∙L–1 2.3 mol∙L–1 2.3 mol∙L–1

— 5 5 5 5 5 5

0.15 0.15 0.15 0.15 0.15 0.15 0.15

Ratio of activator solution weight to solid mixture weight.

Fig. 1—Particle distribution and features of PS. 174

ACI Materials Journal/March-April 2016

Fig. 2—Pisha sandstone-based mortar specimens. stress development between the specimen end and surfaces and polished stainless steel end platens of the loading frame. Mercury intrusion porosimetry (MIP)—The material’s porous space was characterized by mercury intrusion porosimetry (MIP) testing. This technique provides the cumulative pore volume as a function of applied pressure based on the mercury intrusion under increasing pressures. The measurements were carried out using an automated mercury porosimeter over the pressure range between 0.001 and 400 MPa (0.01 bar and 4000 bar). Before testing, specimens were cleaned in a microwave bath and were dried at 80°C (176°F) until constant weight was achieved. Water permeability—The first step in the water permeability test was to cut a 40 mm (1.57 in.) thick slice from the middle of a 100 x 200 mm (3.94 x 7.87 in.) cylinder specimen. The circumference of the slice was covered with 25 mm (0.98 in.) of epoxy resin that was allowed to harden for 24 hours. The specimen was placed in a permeability housing cell, as shown in Fig. 3, and water pressure of 0.5 MPa (5.0 bar) was applied to the cell. This pressure was recommended and used by Chan and Wu14 and Chindaprasirt et al.15 in their research. The amount of water flowing through the specimen was measured by reading the reduction of the water level in a manometer tube. The results were plotted as a graph of the cumulative amount of water flowing as a function of the cumulative time to determine the coefficient of permeability using Darcy’s law and the equation of continuity16

K = ρLgQ/PA (1)

where K is the coefficient of water permeability (m/s); ρ is density of water (kg/m3); g is acceleration due to gravity, 9.8 (m/s2); Q is flow rate (m3/s); L is thickness of the specimen (m); P is water pressure (Pa); and A is the cross-sectional area of the specimen (m2). Fourier transform infrared—Fourier transform infrared (FTIR) analysis was performed using the KBr pellet method (1 mg sample per 100 mg KBr) on a spectrometer, with 32 scans per sample collected from 4000 to 400 cm–1 at 4 cm–1 resolution. Thermogravimetric analysis (TG)—A simultaneous thermal analyzer was used to measure some physical properties of the material as a function of the temperature change. The samples were heated from 50 to 1000°C (122 to 1832°F) in an atmosphere of nitrogen with a heating rate of 10°C·min–1 (50°F·min–1). ACI Materials Journal/March-April 2016

Fig. 3—Experimental setup for testing water permeability of Pisha sandstone-based mortar. Scanning electron microscopy—The microstructure of the cured mortar was examined using a scanning electron microscope (SEM) at an accelerating voltage of 20 kV. Chemical elemental analyses were also performed by an energydispersive X-ray spectroscope device equipped with this SEM system. The samples were fractured to expose the fresh surface before mounting them on aluminum stubs using carbon paint. The samples were then sputter-coated with gold palladium for SEM examination. RESULTS AND DISCUSSION Mechanical properties The effect of activator solution concentration on the compressive strength of PSM and PSFM for different ages is shown in Fig. 4. It can be seen that the NaOH concentration was notably influenced by the properties of PSM. The mechanical strength of PSM was enhanced after alkali activation when compared to PS. The reason for this increase can be concluded that the cementitious materials of PS, carbonate mineral, and clay mineral1 were easy to dissolve in water. For PSM, the binder was geopolymer gel, which was the main contributor of strength. Figure 4(a) shows the relationship between NaOH concentration and compressive strength. The results of Fig. 4(a) showed that the compressive strength has a slight increase with the concentration of the activator solution at an early age (within 28 days); the reason for this could be assumed that the reaction of PS and alkali activator is insufficient at the early stage and the excessive alkali would make a negative effect on the mechanical strength17 due to the excessive free OH– were remaining in the samples, weakening the structure of pastes.18-20 However, the strength has a significant gain with the increasing NaOH concentration at 90 days. It could be ascribed to the excessive free OH– transformed into calcium silicate hydrate gradually along the curing age, and the strength of mortar also increased due to the amount of geopolymer gel increased.21 Figure 4(b) shows the relationship between compressive strength and fly ash dosage of PSFM for different ages. 175

Fig. 4—Compressive strength of Pisha sandstone-based mortar with different NaOH concentration for different ages (N-NaOH). Table 3—MIP results for PSM and PSFM Fly ash content, wt.%

NaOH concentration mol∙L–1

Total porosity, %

PS

—

27.60

2.39 (347)

0

0

1.35

28.20

6.20 (899)

0.62

0

2.3

29.0

7.80 (1131)

0.58

0

3.10

29.10

10.30 (1494)

0.66

5

2.30

27.10

11.30 (1638)

0.82

10

2.30

24.60

15.01 (2176)

0.81

15

2.30

22.50

19.32 (2801)

0.86

According to Fig. 4(b), it was shown that the fly ash dosage plays an important role in the gain of compressive strength. The mechanical strength of specimens would obtain a significant increase when the fly ash was added. When the amounts of fly ash were 5, 10, and 15%, the compressive strength would be increased to 145, 192, and 248%, respectively. Fly ash contains sufficient amounts of reactive alumina and silica, which could be transformed into geopolymer gel when in contact with alkali activator22; therefore, the amount of geopolymer gel increased with the increasing fly ash dosage. The alkali activation product, geopolymer gel, is the main contributor to the strength of PSFM. On the other hand, the addition of fly ash could enhance the density of the matrix, and the pore’s structure would be improved to reduce the porosity of the PSFM as the increasing fly ash dosage, and the effect of porosity on compressive strength would be analyzed in the pore structure section. Pore structure The results of MIP tests in terms of total porosity are given in Table 3 and pore diameter distributions are given in Fig. 5(a) through 5(e). To present a pore size distribution, the log differential distribution curve has the advantage of showing a spectrum of pore size and is especially revealing when the sample has two or more unique peak pore sizes. Figures 5(a) through 5(e) presented the results of MIP tests of PS, PSM, and PSFM. It was observed that the porous volume was generally detected in the pore range of 10 to 30,000 nm (3.94 × 10–7 to 1.18 × 10–3 in), and the pores in PS are subdivided into large pores, which have a diameter of approximately 2090 nm (8.23 × 10–5 in.), and small pores, which have a diameter of 26 nm (1.02 × 10–6 in.) (Fig. 5(a)). The formation of geopolymer gel in the mixtures had signif176

Compressive strength, MPa (psi)

Softening coefficient

icant influence on the pore size distribution; the geopolymer gel will invariably lead to a change in volume, and this change in volume generally gives rise to a large change in porosity and pore size distribution of the samples. Figures 5(b) and 5(c) showed the impact of the NaOH concentration on pore structure of PSM. For PSM samples, the alkali activated had increased the proportion of small pores (10 to 50 nm [3.94 × 10–7 to 1.91 × 10–6 in.]) and reduced the proportion of large pores (1000 to 30,000 nm [3.94 × 10–5 to 1.18 × 10–3 in.]) when compared to the pore distribution of PS. A remarkable increase in the intensity of peak of log differential intrusion corresponding to small pores and a decline in the intensity of peak of log differential intrusion corresponding to large pores were observed after the alkali activated. The proportion of small pores increased with the increasing NaOH concentration; however, the peak of log differential intrusion corresponding to large pores had a shift from 11,309 to 17,278 nm (4.46 × 10–4 to 6.81 × 10–4 in.), the reason for this change can be ascribed to the presence of calcium carbonation, which was the product of the carbonation of C-S-H gel, and the change of the degree of carbonation of C-S-H gel with increasing NaOH concentration would be analysis in TG/DSC. The addition of fly ash to PS in paste leads to the formation of finer and discontinuous pores or to increase in the fraction of finer pores. Figures 5(d) and 5(e) showed the significant influence of fly ash dosage on the decrease of the amount of large pores and the pore size distribution in PSFM samples. The intensity of peak of log differential intrusion corresponded to macropores reduced with the increasing dosage of fly ash. The maximum volume intruded corresponded to a pore diameter of approximately 11 nm (4.33 × 10–7 in.), samples showed a wider distribution with a shift of larger pores in the pore range of 10 to 100 nm (3.94 × 10–7 to 3.94 × 10–6 in.). ACI Materials Journal/March-April 2016

Fig. 5—Pore diameter distributions. (a) PS; (b) N–1.35 fly ash 0%; (c) N-3.1 fly ash 0%; (d) N-2.3 fly ash 10%; and (e) N-2.3 fly ash 15%. Figure 6 shows the relationship between differential intrusion volume and pore size. It can be seen that the large pores’ volume and the total porosity had a slight increase with the NaOH concentration due to the carbonation of C-S-H gel. The addition of fly ash could enhance the density of the matrix, and the pores’ structure would be improved to reduce the porosity of the PSFM as the dosage of fly ash increases; the results of Fig. 6 and Table 3 showed that the proportion of small pores (<50 nm [1.97 × 10–6 in.]) had a remarkable increase and the total porosity decreased from 27.1 to 22.5% with the increasing fly ash content. The porosity and pore structure have a significant influence on the strength and durability of mortar.23 The softening coefficient characterizes materials’ stability to water; it is a key index for evaluating the material durability and water resistance, and the softening coefficient is a ratio of wet compressive strength and dry compressive strength.24 The pore size distribution, the shape, and the position of pores are also important, but it is both difficult and impractical to include all these parameters; it is believed that there is an acceptable relationship between strength and porosity. Figure 7 shows the relationship among strength, softening coefficient, and porosity of ACI Materials Journal/March-April 2016

PSFM. It can be seen that with the decreasing of porosity of PSFM, the value of softening coefficient and compressive strength of PSFM increased from 0.58, 7.8 MPa to 0.76, 19.32 MPa (84, 1131 psi to 110, 2801 psi). There is a good liner correlation between the compressive strength and porosity, and the relationship between softening coefficient and porosity has a same trend. This could be explained that the large pores are sites of weaknesses that control the mechanical properties of mortar, the lower porosity, the more dense pastes, and the less water access; therefore, an increase in porosity will lead to a decrease in strength and water resistance of PSFM. There is an inverse correlation between the strength and porosity of mortar; this is consistent with the findings of previous studies.25,26 Water permeability The values of water permeability of PSFM at 28 and 90 days are shown in Fig. 8(a) and 8(b), respectively. At 28 days, the values of water permeability of PSM3.1, PSFM5, PSFM10, and PSFM15 specimen were 21.9 × 10–8, 14.7 × 10–8, 8.3 × 10–8, and 7.4 × 10–8 m/s (71.8 × 10–8, 48.2 × 10–8, 27.2 × 10–8, and 24.3 × 10–8 ft/s), respectively, 177

Fig. 6—Relationship between differential intrusion volume and pore size of Pisha sandstone-based mortar.

Fig.7—Relationship between compressive strength, softening coefficient, and porosity. and this declined to 16.6 × 10–8, 11.2 × 10–8, 6.3 × 10–8, and 5.1 × 10–8 m/s (54.5 × 10–8, 36.7 × 10–8, 20.7 × 10–8, and 16.7 × 10–8 ft/s) at 90 days. Therefore, the PSM specimen had higher water permeability values than PSFM specimens at both testing ages; as expected, water permeability values decreased with fly ash content, and the mixture with 15% fly ash content showed the minimum permeability values. At 28 and 90 days, the water permeability of PSFM specimens were 21.9 × 10–8 and 7.4 × 10–8 m/s (71.8 × 10–8 and 24.3 × 10–8 ft/s), and these declined to 16.6 × 10–8 and 5.1 × 10–8 m/s (54.5 × 10–8 and 16.7 × 10–8 ft/s) when the fly ash content increased from 0 to 15%. The significant increase in PSFM resistance to water permeability was attributed to improved PSFM pore structure, resulting from pozzolanic reactions of fly ash. The relationship between the porosity and water permeability of PSFM at 90 days is illustrated in Fig. 8(b). As the result, as Fig. 8(b) clearly indicates, the values of water permeability of PSFM increased as the porosity increased. On the other hand, the lower porosity of PSFM resulted in lower water permeability values. The low water permeability of PSFM containing fly ash was affected by the pozzolanic reaction and packing effect of small particles, which produced PSFM with a denser matrix and lower permeability. Fourier transform infrared FTIR analyses for PS, PSM, and PSFM for 90 days are shown in Fig. 9(a) and 9(b). For the PS specimen, it shows the main adsorption bands as follows: 450, 468, 520, 1040, and 1639 cm–1 (1772, 1843, 2047, 4094, and 6453 in.–1). The peaks at 450, 468, 520, and 1040 cm–1 (1772, 1843, 178

2047, and 4094 in.–1) are assigned to Si-O-Si bending vibration bond, whereas the Si-O asymmetric stretching vibration bond can be seen at 1040 cm–1 (4094 in.–1), and these stretching modes are related to quartz (SiO2), the broad band at 1639 cm–1 (6453 in.–1) associated with bending vibrations of H-OH bonds, related to water. The PSM with N-1.35, N-2.3, and N-3.1 (Fig. 9(a)) exhibit infrared vibration modes at 1645 cm–1 (6477 in.–1) characterized the spectrum of stretching and deformation vibrations of the OH and H-O-H group from the weakly bound water molecules, which were trapped in the large cavities between the rings of geopolymer products (C-S-H gel). A mode at 1458 cm–1 (5740 in.–1) corresponding to the stretching vibration of O-C-O bonds in the carbonate group (CO32–) was also observed; the other carbonate C-O stretching vibrations were assigned to regions 875 cm−1 (3445 in.–1) and 713 cm−1 (2807 in.–1). Carbonates were present in the samples due to reaction of older C-S-H gel and CO2 with formation of calcium carbonate (CaCO3), consistent with the observation of calcite in the XRD data.2 The mode at 1036 cm–1 (4079 in.–1) is assigned to the asymmetric stretching vibration of Si-O-T bonds, where T is tetrahedral silicon or aluminum.27 This specific frequency is characteristic of silicon or aluminum in the chain structure of calcium silicate hydrate (C-S-H). The shoulder at 875 cm–1 (3445 in.–1) is associated to the asymmetric stretch of AlO4– groups, and the mode at 713 cm–1 (2807 in.–1) corresponds to the bending of Al-O-Si bonds. The results of Fig. 9(a) showed the effect of NaOH concentrations on the samples. The asymmetric stretching vibration of O-C-O bonds of CO32– groups shifted (1470 to 1442 cm–1 [5787 to 5677 in.–1]); a marked increase in the intensity in the vibration was observed when the concentration of NaOH solution increased. This indicates chemical changes in the reaction products formed by alkaline activation of the PS— in particular the decalcification of the C-S-H gel to form calcium carbonates and amorphous silica.28,29 The mode at 875 cm–1 (3445 in.–1) became sharper and had an increase in the intensity was also associated with the formation of calcite. These indicated that high NaOH concentrations could lead to a higher degree of carbonation of C-S-H gel. The Si-O bond at 1036 cm–1 (4079 in.–1) is on the spectra, for the C-S-H gel became stronger and narrower with the increase of NaOH concentrations. An increase in the wavelength of this peak indicates that higher NaOH concentrations could promote hydration reaction of calcium silicate and lower degree of polymerization of C-S-H gel.30 Figure 9(b) presented FTIR ACI Materials Journal/March-April 2016

spectra of PSFM with different fly ash dosage. It can be seen that there were no significant changes in all bands with the increase of fly ash content. Insignificant shifts occurred in the carbonate C-O stretching vibrations (1475, 875, and 713 cm−1 [5807, 3445, and 2807 in.–1]) and asymmetrical Si-O-Si and Al-O-Si stretch (1035 cm−1 [4075 in.–1]). The carbonate C-O stretching vibrations at 875 cm−1 (3445 in.–1) became weaker and the new bands at 856 cm−1 (3370 in.–1) disappeared, which is associated with carbonate C-O stretching vibrations with the increase of fly ash content. It meant that the carbonation rate of C-S-H gel decreased with the increase of fly ash content. Thermogravimetry analysis Figure 10 showed the influence of NaOH concentration and fly ash dosage on the products of PSM and PSFM. The

TG/DSC curves of specimens are presented in Fig. 10 and Fig. 11. On the TG curve for PSM and PSFM, the two steps of weight loss, together with corresponding endothermic DSC peaks, are observed. The relative weight loss at 50 to 105°C (122 to 221°F) was due to the loss of free water and loosely bound water; in the region of 105 to 500°C (221 to 932°F), the weight loss refers to the loss of structural water that is present in the form of –OH sites in geopolymer gel, and the weight loss in the region of 500 to 780°C (932 to 1436°F) is the decarbonation of the calcite into CaO.31,32 For PSM (Fig. 10(a)), in the region of 105 to 500°C (221 to 932°F), the amounts of water loss of C-S-H gel are 5.2, 5.29, and 5.8%, and the amounts of CO2 loss of CaCO3 (500 to 780°C) are 8.75, 9.04, and 10.08% when the NaOH concentrations are 1.35, 2.3, and 3.1 mol∙L–1, respectively. In the region of 105 to 500°C (221 to 932°F), the higher

Fig. 8—(a) Relationship between water permeability and fly ash content; and (b) relationship between water permeability and porosity.

Fig. 9—(a) FTIR of hydration products of PSM for 90 days; and (b) FTIR of hydration products of PSFM for 90 days.

Fig. 10—(a) TG curves of PSM for 90 days; and (b) TG curves of PSFM for 90 days. (Note: (°C × 1.8) + 32 = °F.) ACI Materials Journal/March-April 2016

179

Fig. 11—DSC curves of specimens for 90 days: (a) PSM; and (b) PSFM. (Note: (°C × 1.8) + 32 = °F.)

Fig. 12—SEM micrographs of specimens for 28 days: (a) N–1.35 fly ash 0%; (b) N-3.1 fly ash 0%; (c) N-2.3 fly ash 5%; and (d) N-2.3 fly ash 15%. NaOH concentration, the more water loss, this means that the amount of C-S-H gel of PSM increased with the increasing NaOH concentration; it also can be seen that a higher NaOH concentration would lead to more CO2 loss of CaCO3, which was formed by the carbonation of C-S-H gel. This indicated that the degree of the carbonation of C-S-H gel would increase with the increasing NaOH concentration; these results agree well with the results of FTIR (Fig. 9(a)). For the PSFM specimen (Fig. 10(b)), in the region of 105 to 500°C (221 to 932°F), the amounts of water loss of C-S-H gel increased from 5.14 to 6.22% when the fly ash content increased from 5 to 15%. However, the amounts of CO2 loss of CaCO3 (500 to 780°C [932 to 1436°F]) decreased from 7.49 to 6.05% when the fly ash content increased from 5 to 15%. Thus, it indicated that a high fly ash addition could decrease the degree of the carbonation of C-S-H gel; this is consistent with the results of Fig. 9(b). The DSC data of Fig. 11 showed the weight loss speed of PSM and PSFM. It can be seen that the peak at a tempera180

ture of approximately 780°C (1436°F) became stronger and had shifted to a high temperature with the increasing NaOH concentration and the decreasing fly ash content. It means that the specimen with a higher NaOH concentration or less fly ash addition had a quicker weight loss speed than the sample which had a lower NaOH concentration or more fly ash addition. In addition, the weight loss speed of PSFM was connected to the degree of carbonation of C-S-H gel; the more CaCO3 was produced, the higher weight the loss speed. These findings are consistent with the results of Fig. 10. Microstructural analysis Scanning electron microscopy images of PSM and PSFM for 28 and 90 days are shown in Fig. 12 and 13. The EDS micrograph of PSFM for 90 days was examined to analyze the morphology of reaction products. The elemental concentration is list in Table 4; each given value represents the average of six readings taken adjacent to each other. The analysis of the morphology of PSFM revealed that, in ACI Materials Journal/March-April 2016

Fig. 13—SEM micrographs of specimens for 90 days: (a) N–1.35 fly ash 0%; (b) N-2.3 fly ash 0%; (c) N-3.1 fly ash 0%; (d) N-2.3 fly ash 5%; (e) N-2.3 fly ash 10%; and (f) N-2.3 fly ash 15%. general, the reaction products of PSFM mainly were C-S-H gel, geopolymer gel, and calcium carbonate after their contact with alkali activator. Figure 12 showed the effects of NaOH concentration and fly ash dosage on the micromorphology and hydration products of PSM and PSFM for 28 days. It can be seen that the degree of hydration is low and the matrix of PSM showed a porous structure at early age. The sample with higher NaOH concentration showed higher density; the amount of geopolymer gel increased with the increasing NaOH concentration (Fig. 12(a) and 12(b)). There are some unreacted fly ash particles coexisting in the geopolymer matrix (Fig. 12(c) and 12(d)), and the sample with more fly ash dosage showed that more unreacted fly ash remained. The SEM of PSM and PSFM for 90 days (Fig. 13) showed a great amount of dense and homogenous geopolymer gels. The morphology change of PSM and PSFM indicated that the NaOH concentration and the dosage of fly ash had a significant effect on the microstructure and the carbonation ratio of C-S-H gel. The results of chemical analysis of PSM and PSFM (which, listed in Table 4, showed that the ratio of Ca/Si fluctuated between 1.01 and 1.77 [except Fig. 13(e)], and the ratio of Al/Si ACI Materials Journal/March-April 2016

changed from 0.26 to 0.38) indicated that the main composition of geopolymer gel was C-S-H.31,32 The NaOH concentration had a significant effect on the microstructure and the carbonation ratio of C-S-H gel; it can be seen that the ratio of Ca/Si decreased from 1.27, 1.19, to 1.01 with the increasing NaOH concentration; according to the analysis results of FTIR and TG, the degree of carbonation of C-S-H gel increased with the increasing NaOH concentration. Thus, it believed that a higher degree of carbonation of C-S-H gel would lead to a decrease in the ratio of Ca/Si of C-S-H gel. However, the ratio of Al/Si has an inexplicit change with the increasing NaOH concentration, and the ratio of Al/Si changed between 0.26 and 0.34 when the NaOH concentration increased from 1.35 to 3.1 mol·L–1. Figures 13(d) through 13(f) showed the effect of fly ash dosage on the hydration and carbonation of alkali-activated PSFM. The morphologies of hydration products and carbonation products showed a great amount of geopolymer gels, C-S-H gel, and few calcium carbonates, which made the matrix form a dense structure. The ratio of Ca/Si increased from 1.48 to 1.77 and the ratio of Al/Si decreased from 0.38 to 0.29 with the increasing fly ash dosage. In particular, the ratio of Ca/Si of Fig. 13(e) showed 181

Table 4—Composition of PSM and PSFM by SEM-EDS Elemental composition, wt.% Points Fig. 13(a)

Fig. 13(b)

Fig. 13(c)

Fig. 13(d)

Fig. 13(e)

Fig. 13(f) *

C

O

Ca

Si

Al

Na

Mg

Others

Ca/Si

1

25.67

50.53

10.89

8.36

2.30

1.44

0.30

0.51

1.30

2

22.29

62.59

6.03

4.95

1.27

1.56

0.82

0.49

3

13.07

64.96

9.05

6.99

1.52

1.16

2.86

0.39

1

7.17

55.17

13.55

10.69

5.72

0.34

5.40

2

11.12

52.19

14.71

13.70

4.01

1.64

3

11.88

59.60

12.17

9.82

1.93

1

26.27

51.72

8.27

8.54

2.29

2

17.02

61.60

6.83

6.74

3

19.32

53.90

10.3

1

23.11

52.10

2

19.51

3

18.59

1

Al/Si —

0.28

—

1.22

—

0.27

—

1.29

1.27*

0.22

0.26*

1.96

1.27

—

0.53

—

1.56

1.07

1.07

—

0.29

—

2.26

1.40

0.94

1.24

1.19

0.20

0.34

1.56

0.63

0.72

0.97

—

0.27

—

2.63

1.60

2.88

0.70

1.01

—

0.39

—

9.86

3.07

1.45

1.54

0.56

1.04

1.01

0.31

0.32

11.15

7.94

1.54

1.44

0.80

0.92

1.40

—

0.20

—

46.83

14.94

9.41

6.02

0.96

0.53

1.80

1.59

—

0.64

—

56.12

10.73

7.38

2.16

1.40

2.53

1.09

1.45

1.48

0.29

0.38

24.63

55.65

9.59

7.03

1.26

1.08

0.68

0.08

1.36

—

0.18

—

2

20.48

64.54

13.90

0.43

0.16

0.24

—

0.25

32.33

—

0.37

—

3

21.92

66.21

8.89

1.58

0.48

0.56

0.24

0.12

5.63

13.11

0.30

0.28

1

22.88

62.73

7.11

3.99

1.48

—

1.37

0.44

1.78

—

0.37

—

2

13.02

47.74

20.04

12.75

4.07

—

1.68

0.70

1.57

—

0.32

—

3

11.58

60.29

15.44

7.82

1.31

1.28

0.59

1.69

1.97

1.77

0.17

0.29

Average value of ratio of Ca/Si, Al/Si of Points 1, 2, and 3.

an especially high value; it indicated that the products of Points 2 and 3 of Fig. 13(e) were calcium carbonate, which were the carbide products of C-S-H gel. The addition of fly ash could reduce the large pores and improve the pore structure and make the matrix more dense; the dense structure would decrease the degree of carbonation of C-S-H gel. It meant that high fly ash dosage could promote the synthesis of geopolymer and decrease the carbonation of C-S-H gel. CONCLUSIONS In this study, the development of compressive strength, pore structure, water permeability, and microstructure of PSM and PSFM was studied. The conclusions can be summarized as follows: 1. The significant factors affecting the compressive strength of Pisha sandstone-based mortar are NaOH concentration and fly ash dosage. At early age (within 28 days), NaOH concentration had a slight effect on the compressive strength. As the NaOH concentration increased, the relative strength decreased due to the excessive alkali that would make a negative effect on the mechanical strength. For PSFM, the compressive strength increased with the increasing dosage of fly ash. However, for the strength of specimens at 90 days, both the NaOH concentration and fly ash dosage had a remarkable influence on the compressive strength; the strength of PSFM increased to a maximum value with the increasing NaOH concentration and the fly ash dosage. With all the other parameters remaining constant, when the amounts of fly ash were 5, 10, and 15%, the compressive strength was increased to 145, 192, and 248%, respectively. 2. The formation of geopolymer gel binders in the mixtures has a significant influence on the pore size distribution. The 182

large pore volume and the total porosity had a slight increase with the NaOH concentration due to the carbonation of C-S-H gel. The addition of fly ash could enhance the density of the matrix and the pore structure would be improved to reduce the porosity of the PSFM with the increasing fly ash dosage; the proportion of small pores (<50 nm) had a remarkable increase and the total porosity decreased from 27.1 to 22.5% with the increasing dosage of fly ash. The sample with 15% fly ash dosage exhibited the lowest porosity and finest pore structure. There is an inverse correlation between the strength and proportion of large pores; with the decreasing of porosity of PSFM, the value of softening coefficient and compressive strength of PSFM increased from 0.58, 7.8 MPa to 0.76, 19.32 MPa (84, 1131 psi to 110, 2801 psi). Large pores are sites of weaknesses that control the mechanical properties of mortar. 3. The use of fly ash as an admixture in PSM exhibited a good result in water permeability. Depending on the fly ash content level, the water permeability of PSFM decreased as the dosage of fly ash increased. The minimum permeability values of PSM and PSFM were 16.6 × 10–8 and 5.1 × 10–8 m/s (54.45 × 10–8 ft/s and 16.73 × 10–8 ft/s) at 90 days. The significant increase in PSFM resistance to water permeability was attributed to improved Pisha sandstone-based mortar pore structure resulting from pozzolanic reactions of fly ash. The low water permeability of PSFM containing fly ash was affected by the pozzolanic reaction and packing effect of small particles, which produced PSFM with a denser matrix and lower permeability. 4. FTIR, TG, and SEM/EDS results indicate that the reaction products of Pisha sandstone-based mortar mainly showed a homogenous geopolymer gel. The geopolymer gel was mainly ascribed to amorphous hydrated calcium silicate ACI Materials Journal/March-April 2016

(C-S-H) gel. The FTIR results showed that the Pisha sandstone-based mortar had carbonation; the result of TG indicated that the amounts of C-S-H gel and the degree of carbonation of C-S-H gel would increase with the increasing NaOH concentration and the decreasing dosage of fly ash. SEM/ EDS results showed that the NaOH concentration and fly ash dosage had a significant effect on the ratio of Ca/Si. The ratio of Ca/Si decreased from 1.27 to 1.01 and 1.77 to 1.48 when the NaOH concentration increased from 1.35 to 3.1 mol/L, and the fly ash dosage decreased from 15 to 5%, respectively. AUTHOR BIOS

ACI member Changming Li is a PhD Graduate Student at the Institute of Building Materials, Faculty of Infrastructure Engineering, Dalian University of Technology, Dalian, Liaoning, People’s Republic of China. His research interests include the micromechanics of cementitious materials and the design, characterization, and application of geopolymer composites. Tingting Zhang is a Lecturer at the Institute of Building Materials, Faculty of Infrastructure Engineering, Dalian University of Technology, China. She received her BS from Dalian University of Technology and her MS and PhD from Imperial College London, London, UK. Her research interests include low-carbon cement and concrete, resources from solid wastes, and solidification of radioactive waste. Lijiu Wang is a Professor at the Institute of Building Materials, Faculty of Infrastructure Engineering, Dalian University of Technology, China. His research interests include the micromechanics of cementitious materials; the design, characterization, and application of geopolymer composites; and materials-based development of sustainable infrastructure.

ACKNOWLEDGMENTS

The authors would like to express gratitude for the financial support by the National Key Science & Technology Pillar Program of China (No. 2013BAC05B03), the National Natural Science Foundation of China (Grant No. 51408096), and the Fundamental Research Funds for the Central Universities (DUT15RC(4)22).

REFERENCES

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ACI Materials Journal/March-April 2016

10. Gao, Z. L., and Yang, S. W., “Existing Problems of Silt Arresters on the Loss Plateau,” Bulletin of Soil and Water Conservation, V. 19, No. 6, 1999, pp. 16-19. (in Chinese with English summary) 11. Yang, E. H.; Yang, Y. Z.; and Li, V. C., “Use of High Volumes of Fly Ash to Improve ECC Mechanical Properties and Material Greenness,” ACI Materials Journal, V. 104, No. 6, Nov.-Dec. 2007, pp. 303-311. 12. Reiner, M., and Rens, K., “High-Volume Fly Ash Concrete: Analysis and Application,” Practice Periodical on Structural Design and Construction, ASCE, V. 11, No. 1, 2006, pp. 58-64. doi: 10.1061/ (ASCE)1084-0680(2006)11:1(58) 13. ASTM C39/C39M, “Standard Test Method for Compressive Strength of Cylindrical Concrete Specimens,” ASTM International, West Conshohocken, PA, 2010, 8 pp. 14. Chan, W. W. J., and Wu, C. M. L., “Durability of Concrete with High Cement Replacement,” Cement and Concrete Research, V. 30, No. 6, 2000, pp. 865-879. doi: 10.1016/S0008-8846(00)00253-2 15. Chindaprasirt, P.; Homwuttiwong, S.; and Sirivivatnanon, V., “Influence of Fly Ash Fineness on Strength, Drying Shrinkage and Sulfate Resistance of Blended Cement Mortar,” Cement and Concrete Research, V. 34, No. 7, 2004, pp. 1087-1092. doi: 10.1016/j.cemconres.2003.11.021 16. Weerachart, T., and Chai, J., “Strength, Drying Shrinkage, and Water Permeability of Concrete Incorporating Ground Palm Oil Fuel Ash,” Cement and Concrete Composites, V. 32, No. 9, 2010, pp. 767-774. 17. Temuujin, J., and van Riessen, A., “Effect of Fly Ash Preliminary Calcinations on the Properties of Geopolymer,” Journal of Hazardous Materials, V. 164, No. 2-3, 2009, pp. 634-639. doi: 10.1016/j.jhazmat.2008.08.065 18. Bakharev, T.; Sanjayan, J. G.; and Cheng, Y. B., “Effect of Admixtures on Properties of Alkali-Actived Slag Concrete,” Cement and Concrete Research, V. 30, No. 9, 2000, pp. 1367-1374. doi: 10.1016/ S0008-8846(00)00349-5 19. Hu, M. Y.; Zhu, X. M.; and Long, F. M., “Alkali Activated Fly Ash-Based Geopolymers with Zeolite or Bentonite as Additives,” Cement and Concrete Composites, V. 31, No. 10, 2009, pp. 762-768. doi: 10.1016/j. cemconcomp.2009.07.006 20. Komnitsas, K.; Zaharaki, D.; and Perdikatsis, V., “Effect of Synthesis Parameters on the Compressive Strength of Low-Calcium Ferronickel Slag Inorganic Polymers,” Journal of Hazardous Materials, V. 161, No. 2-3, 2009, pp. 760-768. doi: 10.1016/j.jhazmat.2008.04.055 21. Oner, A.; Akyuz, S.; and Yildiz, R., “An Experimental Study on Strength Development of Concrete Containing Fly Ash and Optimum Usage of Fly Ash in Concrete,” Cement and Concrete Research, V. 35, No. 6, 2005, pp. 1165-1167. doi: 10.1016/j.cemconres.2004.09.031 22. Hassan, K. E.; Cabrera, J. G.; and Maliehe, R. S., “The Effect of Mineral Admixtures on the Properties of High Performance Concrete,” Cement and Concrete Composites, V. 22, No. 4, 2000, pp. 267-271. doi: 10.1016/S0958-9465(00)00031-7 23. Aligizaki, K. K., Pore Structure of Cement-Based Materials: Testing, Interpretation and Requirements, Taylor and Francis, New York, 2006, 32 pp. 24. Guettala, A.; Abibsi, A.; and Houari, H., “Durability Study of Stabilized Earth Concrete under Both Laboratory and Climatic Conditions Exposure,” Construction and Building Materials, V. 20, No. 3, 2006, pp. 119-127. doi: 10.1016/j.conbuildmat.2005.02.001 25. Mai, Y.-W., and Cotterell, B., “Porosity and Mechanical Properties of Cement Mortar,” Cement and Concrete Research, V. 15, No. 6, 1985, pp. 995-1002. doi: 10.1016/0008-8846(85)90090-0 26. Collins, F., and Sanjayan, J. G., “Microcracking and Strength Development of Alkali Activated Slag Concrete,” Cement and Concrete Research, V. 23, No. 4-5, 2001, pp. 345-352. doi: 10.1016/S0958-9465(01)00003-8 27. Bernal, S. A.; de Gutierrez, R. M.; Provis, J. L.; and Rose, V., “Effect of Silicate Modulus and Metakaolin Incorporation on the Carbonation of Alkali Silicate-Activated Slags,” Cement and Concrete Research, V. 40, No. 6, 2010, pp. 898-907. doi: 10.1016/j.cemconres.2010.02.003 28. Puertas, F.; Palacios, M.; and Vázquez, T., “Carbonation Process of Alkali-Activated Slag Mortars,” Journal of Materials Science, V. 41, No. 10, 2006, pp. 3071-3082. doi: 10.1007/s10853-005-1821-2 29. Palacios, M., and Puertas, F., “Effect of Carbonation on Alkali-Activated Slag Paste,” Journal of the American Ceramic Society, V. 89, No. 10, 2006, pp. 3211-3221. doi: 10.1111/j.1551-2916.2006.01214.x 30. Hajimohammadi, A.; Provis, J. L.; and Van Deventer, J. S. J., “The Effect of Silica Availability on the Mechanism of Geopolymerisation,” Cement and Concrete Research, V. 41, No. 3, 2011, pp. 210-216. doi: 10.1016/j.cemconres.2011.02.001 31. Taylor, H. F. W., Cement Chemistry, A.S.o.C. Engineers, ed., second edition, Thomas Telford, London, UK, 1997, 113 pp. 32. Ben Haha, M.; Le Saout, G.; Winnefeld, F.; and Lothenbach, B., “Influence of Activator Type on Hydration Kinetics, Hydrate Assemblage and Microstructural Development of Alkali-Activated Blast-Furnace Slags,” Cement and Concrete Research, V. 41, No. 3, 2011, pp. 301-310. doi: 10.1016/j.cemconres.2010.11.016

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ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M18

Compressive and Time-Dependent Strength of Concrete Masonry Constructed with Type M Mortar and Grouts Containing High Volume of Fly Ash and Slag by Fernando S. Fonseca, Scott M. Watterson, and Kurt Siggard A testing program was conducted to determine whether concrete masonry prisms constructed with Type M mortar and grouts containing high volumes of supplemental cementitious materials (SCMs) could meet minimum masonry compressive strength requirements. Research focused on replacing portland cement (PC) with Class F fly ash and ground-granulated blast-furnace slag (GGBFS) in quantities larger than those currently allowed by technical standards. In addition, the research evaluated the development of the compressive strength of the prisms with time. Thus, specimens were tested at 14, 28, 42, 56, and 90 days. The control prism group contained grout with only PC. In the second prism group, the grout had Class F fly ash replacing PC at rates of 45, 55, and 65% while in the third prism group the grout had Class F fly ash and GGBFS combinations replacing PC at rates of 65, 75, and 85%. The compressive strength of all prisms exceeded the minimum compressive strength requirement of 10.34 MPa (1500 psi) at 28 days, although the 65% fly ash grout mixture itself did not meet the minimum grout compressive strength of 13.79 MPa (2000 psi) at 28 days. A lower estimate of the ultimate strength of grouted prisms constructed with grouts containing high volumes of SCM can be estimated by multiplying the strength measured at 14 days by 1.2 and 1.3 for prisms with binary and ternary grouts, respectively. Keywords: concrete masonry; fly ash; ground-granulated blast-furnace slag (GGBFS); grout; high-volume fly ash; masonry; masonry prisms; strength evolution; supplementary cementitious materials (SCMs).

INTRODUCTION The masonry building code1 allows the results of axial compression tests on masonry prisms to be the basis for determining the design capacity of masonry elements by permitting the compressive strength of the masonry fmʹ to be that of the prisms. Numerous experimental and analytical research programs have been conducted on different aspects of masonry compressive strength. Topics have included the behavior of masonry under concentric2,3 and eccentric loadings4 and the influence of block geometry,5 mortar strength,6 and mortar bedding type7 on the strength of the masonry; in many cases, both hollow8 and grouted9 prisms have been tested. Although grout type affects the compressive strength of masonry,5 different types of grout have not been explored as a means to make masonry construction more competitive and sustainable. Concrete masonry has many proven sustainable benefits and could become even more sustainable if the use of portland cement (PC), whose production generates approximately 1 ton of carbon dioxide per produced ton,10 is reduced. A possible way to achieve such a vision is to ACI Materials Journal/March-April 2016

increase the substitution levels of supplemental cementitious materials (SCMs) for PC in masonry grout. The construction industry has used SCMs to replace ordinary PC in many applications.11 Replacing PC with SCMs have technical12 (that is, durability to chemical attack) as well as economic and environmental benefits.10,13 Two common forms of recycled SCMs are pozzolans and slags.14 Abounding data is available demonstrating the use of SCMs in concrete in typical amounts15 and in high volumes,16 as well as in the manufacturing of masonry units.17 However, limited data exist demonstrating the use of SCMs in masonry grouts.18 A testing program was devised to determine if masonry with grouts containing high levels of SCMs could meet the minimum prescribed fmʹ and to evaluate the development of prism strength with time. The first phase determined the compressive strength of masonry grouts made with various combinations of Class F fly ash (FA) and ground-granulated blast-furnace slag (GGBFS) to replace high amounts of PC.18 The first phase had three stages. In the first and second stages, PC was replaced only by FA, while in the third stage, PC was replaced by combinations of FA and GGBFS. In the first stage, mixtures were proportioned by volume and batched with 0, 20, 30, 40, 50, and 60% PC replacement; specimens were cured in a dry and a wet environment. In the second stage, mixtures were proportioned by weight and batched with 0, 20, 30, 40, 45, 50, 55, 60, and 65% PC replacement. Specimens in the second stage were cured in a wet environment only. In the third stage, mixtures were proportioned also by weight and batched with 0, 50, 60, 65, 70, 75, 80, and 85% PC replacement with the FA content maintained constant at 25%. Specimens in the third stage were also cured in a wet environment only. Grout specimens were tested at 14, 28, 42, 56, and 90 days and three specimens were tested for each replacement rate, age, and curing conditions. Grout mixtures with up to 55% FA and 85% FA-GGBFS substitutions reached the minimum compressive strength of 13.79 MPa (2000 psi) required at 28 days,19 while mixtures with 60 and 65% FA achieved the minimum compressive strength in 44 and 54 days, respectively. The second phase of the testing program determined if masonry prisms with grouts containing high levels of SCMs ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-133.R1, doi: 10.14359/51688638, received May 6, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

185

met the minimum fmʹ of 10.34 MPa (1500 psi) at 28 days.1 The prisms were constructed with Type M mortar and grouts made with various combinations of FA and GGBFS. This second phase of the testing program and corresponding results are presented herein. RESEARCH SIGNIFICANCE The compressive strength of concrete masonry fmʹ, as a result of the prism test method, must be equal to or exceed 10.34 MPa (1500 psi).1 In addition, the code stipulates that the specified compressive strength of the grout fgʹ must be equal to or exceed fmʹ. The code, therefore, by itself, allows fgʹ equal to or greater than 10.34 MPa (1500 psi). The specifications,1 however, require that grout must conform to standards19 that dictate a minimum fgʹ of 13.79 MPa (2000 psi) at 28 days. Had it not been for the specifications, both fmʹ and fgʹ could be equal to 10.34 MPa (1500 psi). Research18 has shown that masonry grout with up to 55% FA and 85% FA-GGBFS substitutions reached the minimum fgʹ of 13.79 MPa (2000 psi), as required by the standards19; grout mixtures with 60 and 65% FA, however, did not reach the minimum fgʹ until 44 and 54 days, respectively. Grout mixture with 60% FA did reach the minimum fmʹ of 10.34 MPa (1500 psi) at 28 days, while grout mixture with 65% FA reached the minimum fmʹ of 10.34 MPa (1500 psi) at 37 days. The research presented herein demonstrates that fmʹ of systems constructed with grouts containing high volumes of SCMs can achieve the minimum fmʹ. Research focused on replacing PC with Class F FA and GGBFS in quantities larger than those currently allowed. This research will help mainstream SCMs as replacement to PC in masonry grout. Furthermore, greater quantities of recycled cementitious materials supplementing the PC content in masonry grouts promotes and brands masonry construction as costand planet-conscious. This research provides engineers with additional means to build sustainable concrete masonry structures by promoting broader SCM addition rates for masonry grout and by extending discretion to engineers in lengthening the minimum curing ages at which grout and masonry strength minimums are achieved. EXPERIMENTAL PROGRAM Concrete masonry units (CMUs), mortar, and grout being components of masonry prisms were tested individually to assure compliance with code and standards and to attribute prism strength gain appropriately. Because concrete materials with pozzolans or slag gain strength over time,11,12 specimens were tested at 14, 28, 42, 56, and 90 days to determine the curing age at which compressive strength minimums were achieved and the time-dependent strength of the masonry. Materials Concrete masonry units—The CMUs used were all from the same batch using consistent fabrication methods and complying with current standards.20 These units were 200 x 200 x 200 mm (8 x 8 x 8 in.) nominal size with one cell. The unit size was selected primarily so that prisms could fit into and not exceed the loading capacity of the testing 186

machine. In addition, the size selections provided easier maneuverability, decreased physical labor in the laboratory environment, and reduced laboratory space consumption. Half units sash blocks were used; a sash block has an end slot to receive metal window frames or premolded expansion joint material. Following standard procedures,21 six units were tested for absorption, moisture content, and density. The average sash square groove dimension was 19.1 mm (0.75 in.) and the average unit height, length, web thickness, and face shell thickness were 193.4 mm, 194.5 mm, 33.5 mm, and 46.2 mm (7.61, 7.66, 1.32, and 1.82 in.), respectively. Although the sash grooves could be neglected,21 researchers accounted for them in determining the CMUs’ area. The average net and gross areas of the CMUs were 24,560 and 37,468 mm2 (38.07 and 58.08 in.2), respectively. The measured and reported average absorptions were 6.60 and 8.76%, respectively; the measured and reported average densities were 2.05 and 1.77 g/cm3 (128 and 111 lb/ft3), respectively; and the measured and reported average moisture content were 43.01 and 56.51%, respectively. The reported values were provided by the manufacturer. Mortar—Commercial-grade Type M mortar was used. The mixture was a dry preblended mixture of sand, cement, and chemical admixtures meeting current standards and requiring only proper amounts of water and mixing for use. Type M mortar is generally the least workable mortar in its plastic state, while in its hardened state, it is generally the strongest in compression and tension. The mortar manufacturer reported a minimum compressive strength of 17.2 MPa (2500 psi). Grout—Fonseca et al.18 gives a detailed description of the grout materials used in this research. The materials used in making the grout complied with appropriate standards, and were PC Type I/II,22 Class F FA,23 Grade 100 GGBFS,24 fine aggregate (sand),25 9.5 mm (0.375 in.) aggregate (pea gravel),25 and potable water; the properties of the materials are given in Fonseca et al.18 Seven variations of grout were used. Standards19 regulate the use of SCMs in masonry grout. Replacement guidelines for SCMs in masonry grout are analogous to limitations of blended hydraulic cement,26 limiting the maximum pozzolan content to 40% by mass of the binary cement and the total content of pozzolan and GGBFS to less than 70% by mass of the ternary cement. An all Type I/II PC-based grout was used as the control mixture. Binary grouts were composed of PC and FA and the PC content was replaced by 45, 55, and 65% FA. Ternary grouts were composed of PC, FA, and GGBFS, and the PC content was replaced with 25% FA and 40, 50, and 60% GGBFS. Specimens and testing The construction and the testing of all specimens were conducted using standard test methods. No special requirement exists for the testing of CMUs,21 except that they are capped.27 The capping was completed with a high-strength gypsum cement compound with 34.5 MPa (5000 psi) compressive strength and, on average, was 3 mm (0.118 in.) thick. Three concrete masonry units ACI Materials Journal/March-April 2016

Fig. 1—Construction of grout specimens.

Fig. 2—Prism construction. were tested for their compressive strength at each age and were surface dried at the time of testing. Professional masons prepared the mortar according to the specifications.1 A mechanical mixer was used and batch sizes were at the discretion of the masons; water amounts were based on the desired workability and consistency. Researchers, however, determined the temperature28 and flow29 of the mortar and prepared the mortar specimens accordingly.28 The average mortar temperature was 21.1°C (70°F) and the flow was 100.4%, a slightly low value because laboratory prepared mortar is expected to have a flow of approximately 110%. Nevertheless, masons approved the mortar and were able to construct the prisms without issue. Mortar specimens were cast according to technical standards.28 Three specimens for each test age were made in 50 mm (2 in.) cube molds. Upon completion of the molding, specimens were placed in a moist room under a plastic sheet to prevent ponding but allowing exposure to the moist air. Specimens remained undisturbed in their molds for 72 hours, after which they were removed from the mold and stored in the moist room until reaching their respective testing ages. The compressive strength determined was outlined in the appropriate standards.28 Specimens were not capped because the final specimens had two near-perfect parallel and smooth surfaces. Grout was made following specific technical standards19 and is described in detail in Fonseca et al.18 Grout should slump19 between 200 and 280 mm (8 and 11 in.) but with the addition of the fine particles making up FA and GGFBS slump was expected to increase as greater percentages of these materials were added to the mixture. Thus, the control mixture was targeted to have a 200 mm (8 in.) slump due ACI Materials Journal/March-April 2016

to foreshadowed increases in slump. The slump test was performed as outlined in appropriate standards.30 Slump was 229, 241, 267, flowable, 203, 216, and 267 mm (9.0, 9.5, 10.5, flowable, 8.0, 8.5, and 10.5 in.) for the control, binary, and ternary grout mixtures, respectively; the 65 binary grout mixture slumped more than 280 mm (11 in.). Five specimens of each grout type for each age were made according to technical standards.31 The specimens were made by filling the cells of CMU blocks—an alternative method allowed by the standards. The CMU cells were filled in two layers of approximately equal depth. The grout specimens and filling of the CMU cells are shown in Fig. 1. The grout specimens were saw cut from the CMU and placed in a moist room of 100% humidity at 31°C (87.8°F) until the day of testing. Grout specimens were capped as per technical standards32 with a high-strength gypsum cement compound and, on average, the capping thickness was 3 mm (0.118 in.). Professional masons assembled the masonry prisms with the aforementioned components. Prisms were constructed in an open, moisture-tight bag large enough to enclose and seal the completed prism on a flat and level base.33 Prism construction is shown in Fig. 2(a). Units were laid in stack bond with full mortar beds and were free of surface moisture at the time of construction. The masons were instructed to assemble the prisms with a 10 mm (0.39 in.) mortar joint thickness; this distance, however, varied slightly and the joint was, on average, 12.0 mm (0.47 in.) thick. Mortar thickness plays a role in the strength of prisms, especially in ungrouted prisms, but for grouted prisms, the effects of the thickness are reduced due to the continuity of the grout.34 Four prisms for each test age and each grout type were constructed. The grouts used in the prisms were of the same 187

batch as those used in the construction of the grout specimens. While waiting to be grouted, prisms were sealed in the moisture-tight bag. Prior to grouting, mortar fins and droppings were removed from inside the prisms. Prisms were grouted between 24 and 48 hours following construction of the prism. Grout consolidation procedures, representative of those used in construction, were carried out with a low-force vibrator, where additional grout was placed into the prisms after consolidation. To have a more uniform solidly grouted prism, reconsolidation and grout topping-off was necessary as the CMUs absorbed water from the grout. Following grouting, prisms were resealed in the moisture-tight bag. Prisms remained in the moisture-tight bags until 48 hours before their respective test age and stayed in their construction location for 48 hours. After 48 hours, prisms were moved a short distance to a lesser traveled area of the laboratory where floor space consumption and possible disturbance would be minimized. The prism construction and storage locations are shown in Fig. 2(b) and 2(c), respectively. Prisms that had CMUs significantly misaligned or gaps between the mortar and CMUs were discarded. Some prisms had protrusions of mortar and grout from construction, which were removed from the top and bottom surfaces with an abrasive stone. Capping of the prisms was completed as required by the governing standards33 using a high-strength gypsum cement compound with 34.5 MPa (5000 psi) compressive strength. The average cap thickness was 3 mm (0.118 in.). Standards required the compressive strength to be adjusted based on the prism height-to-thickness ratio.33 Just prior to testing, prism height was measured and the average heightto-thickness ratio was 2.06, which fell between accepted 1.3 to 5.0 ratios.33 A correction factor of 1.0 is to be used for a 2.0 ratio and 1.04 for a 2.5 ratio. The linearly interpolated correction factor for the 2.06 ratio of the prisms tested yielded a value of 1.0048, which was taken as 1.0. All specimens were tested on a compression testing machine. The upper and lower platens were spherically seated and steel plates of sufficient thickness, according to the dimension requirements of the governing standard,33 were attached to the upper and bottom platens. While half of the expected load could be applied to the prism specimens at any rate and the latter half should be completed at a uniform rate taking between 1 and 2 minutes, a continuous strain rate was applied from initial loading until failure; the testing strain rate was 1.27 mm/min (0.05 in./min) The load was applied until a fracture pattern was visible and the load had significantly decreased in value. In some cases, prisms did not show enough external physical characteristics to determine a mode of failure. In such cases, loading continued until enough visual evidence of a mode of failure was present. RESULTS AND DISCUSSION Time-dependent properties There are several models to represent the time-dependent compressive strength of cementitious materials.12,35 The development of a new model, comparison of models, and an exhaustive discussion of time-dependent properties of the 188

Table 1—Concrete masonry unit results Age, days

ƒcmu, MPa

ƒʹcmu, MPa

COV, %

14

28.0, 24.6, 23.5

25.4

9.2

28

21.2, 20.9, 27.9

23.3

17.0

42

26.5, 26.6, 25.7

26.3

1.8

56

26.1, 24.0, 30.1, 26.1

26.6

9.7

90

28.3, 28.8, 27.6, 26.3

27.7

3.9

Note: 1 MPa = 0.145 ksi.

individual materials used in the construction of the prisms as well as of the prisms are beyond the scope of this paper. The data presented herein are from 14 to 90 days and simple linear regression models are used to show time-dependent strength trends. CMU compressive strength Compressive strength results for the CMUs are presented in Table 1. Three specimens were tested at 14, 28, and 42 days and four specimens were tested for at 56 and 90 days. Coefficients of variation were less than 10% for all testing ages except for 28 days. Inexplicably, two specimens had low compressive strength values, while one specimen had a high compressive strength value causing the mean strength at 28 days to be lower than that at 14 days, which is an anomaly because strength should increase with age. No visual differences were apparent among the specimens but it is plausible that capping may have been at fault. Another possibility is a slightly larger misalignment between the longitudinal axis of the CMUs and that of the machine for some of the specimens. Visual examination after testing indicated that all specimens in the 28-day testing failed due to the crushing of only the face placed away from the machine operator, as shown in Fig. 3. Although the machine swivel upper and lower steel platens should have corrected occurring misalignment, it is plausible that the misalignment was slightly larger, causing an eccentric load to be applied to those specimens. The mean compressive strengths as well as the linear regression model for the CMUs are shown in Fig. 4. The coefficient of determination R2 is low because the mean (calculated) compressive strength value at 28 days is lower than what it should have been because strength should increase with age. The CMUs used in this research were manufactured specifically for this testing program and were received within approximately 10 days from the manufacturing date. Thus, researchers expected a slight increase in CMU strength over time as depicted by the linear model. Using the simple linear model, the CMU strengths at 14 and 90 days are approximately 24.5 and 27.8 MPa (3556 and 4035 psi), respectively; the increase in strength is approximately 13.5% over the period. Mortar compressive strength Compressive strength results for the mortar are presented in Table 2; three specimens were tested for all ages. Coefficients of variation were higher than that for the CMUs, which was expected; nevertheless, the coefficients were less than 20% for all testing ages. The mean compressive strengths as well as the linear regression model response ACI Materials Journal/March-April 2016

Fig. 3—Failure of CMUs at 28-day testing. Table 2—Mortar results Age, days

ƒmortar, MPa

ƒʹmortar, MPa

COV, %

14

21.4, 16.6, 15.2

17.7

18.4

28

21.0, 20.0, 15.5

18.8

15.4

42

23.9, 26.0, 19.8

23.2

13.6

56

19.7, 20.1, 14.5

18.1

17.2

90

20.5, 25.7, 18.0

21.4

18.4

Note: 1 MPa = 0.145 ksi.

is shown in Fig. 5. The strength was expected to increase gradually with age but the results are slightly erratic in that the mean strength at 42 days appears to be slightly higher than the mean strengths at subsequent ages, while the mean strength at 56 days appears to be slightly lower than the mean strengths at earlier ages. The consequence of these discrepancies is that the coefficient of determination R2 of the model is low. The model gives an increase in mortar strength for the period of approximately 14.5%.

ACI Materials Journal/March-April 2016

Fig. 4—Compressive strength of CMUs. Grout compressive strength Fonseca et al.18 summarize and discuss the results of the grout experimental program. Because the grout is one the main variables of this research, however, a brief summary is presented herein with a simple discussion on the time-dependent strength of the grout, which has not been discussed in Fonseca et al.18 Because an alternative method of forming the grout specimens was used, a conversion factor31 based on comparative testing of standard specimens 189

Fig. 5—Compressive strength of mortar. was used to correct the results from the alternative method. To determine the correlation factor, 12 specimens were cast using the standard specimen mold and 12 specimens were cast using the alternative mold. The specimens cast using the alternative method were removed from the CMUs using the same procedure used previously. These 24 grout specimens were then tested using the same procedure previously used and the correlation factor was determined to be 1.2. The results obtained using the alternative method and presented herein have been appropriately corrected. The average compressive strength values for the control grout mixtures and binary grout mixtures are shown in Fig. 6(a); the straight lines connecting the data points are shown for easier identification of the data. At early ages, the 45B and 55B grout mixtures gained strength at a slightly greater rate than the other specimens. After 42 days, a slight decrease is the strength gain rate is observed for both grout mixtures; the decrease for the 55B grout mixture was slightly more pronounced. After 42 days, the strength gain for the 55B and 65B grout mixtures was similar to that of the control mixture, while the strength gain for the 45B grout mixture indicates that it might ultimately become stronger than the control mixture. The data also show that the 65B grout mixture did not reach the minimum compressive strength at the prescribed 28 days. The strength of the mixture at 28 days was 8.29 MPa (1200 psi) but it reached the minimum compressive strength at the age of 54 days. Using the straight lines connecting the data points as reference, Fig. 6(a) shows a slope change for the binary mixture at the age of 42 days. The slope change is probably due to the different rates of strength gain of the hydration products. The main hydration product of PC is a calcium silicate hydrate (C-S-H). In addition to C-S-H, the hydration process also forms calcium hydroxide (CH).11,36 C-S-H and CH are the basis for the effectiveness of the FA. The FA adds strength to the material by means of the FA secondary reacting with CH11,12,37 to form additional C-S-H, which only occurs after the primary reaction has produced enough CH; the secondary reaction is slower than the primary reaction. Thus, the initial gain in strength comes from the primary reaction, which is almost complete at approximately 42 days. Little increase in strength would come from the primary reaction after the 42 days; in other words, the strength curve would be approximately flat (zero slope) after 42 days. Due, to the secondary reaction of FA with the CH, however, the strength starts to increase but at a slower rate than that of the primary reaction. 190

Fig. 6—Compressive strength of binary grout mixtures. The linear regression models representing the binary grouts are shown in Fig. 6(b). It appears that the ultimate strength of the 45B grout will surpass that of the control grout, which is consistent with other research findings.11 The ultimate strength of the 55B and 65B grouts will not reach that of the control grout because of the higher replacement rate of PC. The linear models indicate that from 14 to 90 days the strengths increased 42, 96, 175, and 280% for the control, 45B, 55B, and 65B mixtures, respectively; these increases are significant. The average compressive strength values for the control grout mixture and for the ternary grout mixtures are shown in Fig. 7(a); the straight lines connecting the data points are shown for easier identification of the data. The results for the 75T grout mixture have, at 28 days, what appears to be a discrepancy because the measured strength is smaller than that at 14 days. Faulty caps, misalignment of the end plates, FA and GGBFS flocculation, or a combination thereof may explain the discrepancy. Also noticeable is the significant increase in strength from 28 to 42 days for the 65T and 75T grout mixtures. The large increase for the 75T grout mixture may not be realistic because it is due to the possible erroneous 28-day value. Although the results show a large reduction in strength for the 85T grout mixture, when compared with the strength of the control grout mixture, the mixture did reach the minimum compressive strength of 13.79 MPa (2000 psi) at the prescribed 28 days. Using the straight lines connecting the data points as reference, Fig. 7(a) shows a slope change for the 65T grout mixture at the age of 42 days. If the 28-day value for the 75T grout ACI Materials Journal/March-April 2016

Fig. 7—Compressive strength of ternary grout mixtures. mixture is neglected (because it may be wrong), both the 75T and 85T grout mixtures experienced a gradual increase in strength. Unlike fly ash, GGBFS will hydrate directly to form C-S-H. This process is very slow unless the GGBFS is activated by the alkaline compound CH,11,12,38 which is fortunately formed during the PC hydration. The noticeable change in slope in the strength curve for the 65T grout mixture at 42 days is similar to that of the strength curve for the binary mixtures and is possibly due to the additional initial strength gain from the activated GGBFS; as the reaction slows down, the strength gain rate decreases (and the slope decreases). The 75T and 85T grout mixtures gain strength gradually because the activated GGBFS does not augment the strength but instead simply replaces the PC strength lost due to the large amount of PC that has been replaced. The linear regression models representing the ternary grouts are shown in Fig. 7(b). Compared to mixtures with no cement replacement, mixtures incorporating GGBFS have slightly slower compressive strength development but may have equivalent or even higher ultimate strength. The linear models indicate that the strength gain from 14 to 90 days are approximately 97, 72, and 81% for the 65T, 75T, and 85T grout mixtures, respectively. Although the increases are not as large as those experienced by the binary grout mixtures, they are still significant. Prism compressive strength Compressive strength results for the prisms are presented in Table 3; the mean compressive strengths, with the coefACI Materials Journal/March-April 2016

ficients of variation in parentheses, are also given. Three prisms were tested at 14 days for all replacement rates and four prisms were tested at 28, 42, 56, and 90 days for all replacement rates except for the 65% FA-GGBFS replacement rate, which had only three prisms tested. Considering the fact that only three or four specimens were tested, the variation is very small, giving some reliability and robustness to the obtained results. In general, strength of cementitious materials should increase with age. Thus, the following (small) discrepancies in average compressive strength for the different prism groups are noted. • Control group: at 28 days (it appears erroneous because it is smaller than that at 14 days) and at 90 days (slightly smaller than that at 56 days) • Prisms with 45B grout: at 28 days (slightly lower compared to the values at 14 and 42 days) • Prisms with 55B grout: at 56 days (slightly lower compared to the values at 42 and 90 days) • Prisms with 65B grout: at 90 days (slightly lower compared to the value at 56 days) • Prisms with 65T grout: at 56 days (slightly lower compared to the values at 42 and 90 days) • Prisms with 85T grout: at 28 days (it appears erroneous because it is smaller than that at 14 days) The slopes, y-intercepts, and coefficient of determination, R2, for the linear models are presented in Table 4. The relative low R2 values may be explained by the aforementioned small discrepancies, especially for the control group, due to the possibility that the value at 28 days may be erroneous. The average compressive strength values, as well as the linear regression models representing the strength curves of the prisms, are shown in Fig. 8 for the prisms with binary grouts and Fig. 9 for the prisms with the ternary grouts. The strength evolution of the CMUs, mortar, and grout control group are also shown, in the background, for overall comparison. The compressive strength gain of the prisms generally follows that of the CMUs and mortar but does not resemble that of the grouts except slightly that of the control grout mixture. The overall strength of the prisms with the control grout mixture is smaller than that of the actual grout control mixture, slightly smaller than that of the CMUs, and slightly greater than that of the mortar. Several factors cause a lower strength of grouted prisms, including the initial tension due to restrained drying shrinkage of the grout, the effects of gaps due to incomplete grout compaction, and grout plastic shrinkage. In addition, the incompatibility between the stress-strain properties of the grout and those of the block can cause lateral forces on the block, resulting in earlier failure of the system. Furthermore, the tapered or flared shape of the face shells and webs can result in the grout acting as a wedge and causing earlier failure of the system.39 Although mortar also influences the behavior and strength of prisms, tests results2 indicate that for a reasonable range, mortar strength has no appreciable effect on compressive strength of hollow and grouted prisms—one of the reasons being the continuity of the grout.34 The mortar thickness of the prisms, however, varied slightly and the joint was, on average, 12.0 mm 191

Table 3—Prisms individual results ƒm, MPa

Age, days

Sample

Control

45PB

55PB

65PB

65PT

75PT

85PT

1

20.8

17.4

15.1

14.1

20.2

17.0

18.5

2

23.1

16.7

11.8

14.5

21.1

19.7

18.2

3

22.8

16.0

17.6

9.8

22.3

19.9

17.5

Mean

22.2 (5.6)

16.7 (4.3)

14.8 (19.6)

12.8 (20.7)

21.2 (5.0)

18.9 (8.6)

18.1 (2.7)

1

21.2

16.2

15.7

15.6

24.3

18.1

15.6

2

19.2

15.5

16.7

15.4

23.8

19.0

16.5

3

19.5

17.0

14.9

13.6

19.7

19.2

15.6

4

20.5

17.6

15.2

13.3

20.7

17.8

15.3

Mean

20.1 (4.5)

16.6 (5.4)

15.6 (5.2)

14.5 (8.4)

22.1 (10.3)

18.5 (3.6)

15.7 (3.1)

1

24.0

18.7

18.8

17.0

25.5

25.3

21.6

2

25.8

20.8

18.6

13.0

27.3

22.8

19.8

3

23.4

22.0

18.0

18.9

28.1

23.3

15.4

4

26.7

20.7

16.9

13.3

—

24.9

21.1

Mean

25.0 (6.1)

20.6 (6.6)

18.1 (4.8)

15.5 (18.4)

27.0 (4.9)

24.1 (5.1)

19.5 (14.4)

1

26.3

19.0

16.3

17.0

25.3

25.5

21.8

2

29.0

23.3

16.1

19.0

24.3

22.4

21.0

3

25.0

20.4

18.3

19.6

26.0

20.4

23.2

4

24.9

21.3

19.8

18.5

26.8

24.1

23.8

Mean

26.3 (7.1)

21.0 (8.6)

17.6 (10.0)

18.5 (6.0)

25.6 (4.1)

23.1 (9.6)

22.5 (5.8)

1

24.6

22.0

22.4

20.7

29.5

23.7

23.3

2

24.8

25.8

22.3

15.2

27.2

25.8

23.8

3

25.7

24.8

22.5

18.4

27.9

25.7

24.2

14

28

42

56

90

4

27.4

24.6

20.8

20.3

29.6

25.0

25.2

Mean

25.6 (5.1)

24.3 (6.7)

22.0 (3.7)

18.6 (13.4)

28.6 (4.2)

25.1 (3.8)

24.1 (3.4)

Notes: PB is prisms with binary grout mixtures; PT is prisms with ternary grout mixtures; 1 MPa = 0.145 ksi.

Table 4—Regression and determination coefficients for prisms Coefficient

Control

45PB

55PB

65PB

65PT

75PT

85PT

Slope

0.0645

0.1071

0.0931

0.0803

0.096

0.0883

0.1018

Intercept

20.867

14.89

13.347

12.298

20.453

17.855

15.299

R

0.5177

0.9231

0.9448

0.8413

0.7997

0.7191

0.7817

2

Notes: PB is prisms with binary grout mixtures; PT is prisms with ternary grout mixtures.

(0.47 in.) thick rather than the 10 mm (0.39 in.) typical and expected value. Thus, it is plausible that the strength of the prisms may have been slightly reduced due to the thicker mortar joint. As shown in Fig. 8, the evolution trend of the compressive strength of all prisms with binary grouts follows that of the control prisms. In addition, the results indicate that the overall strength of the prisms with binary grouts is increasingly smaller than that of the control prims; in other words, as the percentage of FA increases, the compressive strength of the binary prisms decreases. The observed reduction is an indication that the strength of the grout does influence the strength of the masonry. Using the linear models, the decreases in strength from the control group at 28 days are approximately 25, 33, and 39% for the 45PB, 55PB, and 65PB prism sets, respectively. At 90 days, the decreases 192

Fig. 8—Compressive strength of prisms with binary grouts.

ACI Materials Journal/March-April 2016

are approximately 8, 19, and 27%, respectively. The linear models indicate that from 14 to 90 days, the strengths increased approximately 23, 33, 48, and 46% for the

control, 45PB, 55PB, and 65PB prism sets, respectively. Time-dependent increase in strength for the prisms with binary grouts are not as pronounced as the increases experienced by the binary grouts themselves, which were 96, 175, and 280%, respectively. As shown in Fig. 9, the evolution trend of the compressive strength of all prisms with ternary grouts also follows that of the control prisms. The 65PT prisms experienced a slight increase in strength relative to that of the control prisms, while there is a decrease in strength for the other ternary prisms as the percentage of FA-GGBFS combination increases; the decrease, however, is not as pronounced as that observed for the prisms with only FA replacement. The 65PT prisms experienced approximately a 3% decrease and 9% increase in strength at 28 and 90 days, respectively. The 75PT and 85PT prisms experienced approximately a decrease in strength at 28 days of 16 and 26%, respectively, and at 90 days a decrease of approximately 3 and 8%,

Fig. 9—Compressive strength of prisms with ternary grouts.

Fig. 10—Failure of prisms. Table 5—Failure modes of prisms Days 14

28

42

56

90

Specimen

Control

45PB

55PB

65PB

65PT

75PT

85PT

1

7

1

3

—

—

4

2

2

5

3

2

3

3

5

3

3

3

2

3

2

5

4

2

1

2

1

3

1

6

3

3

2

3

1

2

1

6

3

2

3

3

1

2

3

5

2

3

4

1

1

2

1

3

3

2

1

5

3

3

3

2

5

2

2

1

2

2

1

5

5

3

3

3

5

2

5

5

2

5

4

5

5

3

2

—

5

5

1

3

1

2

2

1

5

2

2

1

5

3

2

3

5

2

3

1

2

5

5

1

2

1

4

1

3

3

3

2

5, 6

1

1

2

2

5

5

2, 5

5, 6

3, 6

2

2

5

2

6

1

5, 6

2, 5

3

2

5, 6

2

6

1, 5

2, 6

2, 5

4

5

5, 6

6

3

1, 6

3

1

Notes: PB is prisms with binary grout mixtures; PT is prisms with ternary grout mixtures; (1) is conical break; (2) is cone and shear; (3) is cone and split; (4) is tension break; (5) is semi-conical break; (6) is shear break; and (7) is face shell separations.

ACI Materials Journal/March-April 2016

193

respectively. The strengths of the prisms with ternary grouts increased from 14 to 90 days approximately 33, 36, and 47% for the 65PT, 75PT, and 85PT prism sets, respectively; these increases are very similar to those experienced by the prisms with binary grouts. The time-dependent increase in strength for the prisms with ternary grouts are also not as noticeable as the increase experienced by the ternary grouts, which were 97, 72, and 81%, respectively. Although most prisms experienced a decrease in strength relative to the control prisms, all prisms reached the minimum fmʹ of 10.34 MPa (1500 psi) at 28 days. Prism failure The mode of failure of all prisms were recorded as prescribed by technical standards.33 Figure 10 shows the failure of some of the prisms. Table 5 presents the failure mode of each specimen using the following shorthand numerical designation: 1) for conical break; 2) for cone and shear; 3) for cone and split; 4) for tension break; 5) for semi-conical break; 6) for shear break; and 7) for face shell separations. The prisms were loaded until failure, after which the mode of failure was recorded. In some cases, assessing the exact mode of failure was difficult and only the final mode of failure was assessed. Also, distinguishing between one mode of failure and another was difficult; in these cases, both modes are reported. The dual mode of failure simply indicates that the fracture pattern on one mode was obscured by the fracture pattern of the other mode making the distinction between them possible; it is probable that either one is the actual failure mode. Approximately 16% of the prisms failed due to a conical break, 26% experienced a cone and shear failure, 22% experienced a cone and split failure; 1% failed due to a tension break, 24% experienced a semi-conical break, 9% experienced shear break, and 1% experienced face shell separation. Mistakenly, the record for two prisms was lost; these account for less than 1%. CONCLUSIONS The research presented herein provides engineers with additional means to build sustainable concrete masonry structures by promoting broader SCM addition rates for masonry grout. To achieve the objective of the research, prisms were constructed with Type M mortar and seven variations of grout with high levels of SCMs and tested to determine if the prisms could meet the minimum fmʹ of 10.34 MPa (1500 psi) at 28 days. The control prism group contained grout with only PC; the binary prism group had grout with FA replacing PC at 45, 55, and 65%; and the ternary prism group had grout with FA and GGBFS combinations replacing PC at 65, 75, and 85%. The compressive strength of the prisms was determined at 14, 28, 42, 56, and 90 days. From the study conducted, the following conclusions can be made: 1. Masonry prisms constructed with Type M mortar and grouts with up to 65% FA or grouts with up to 25% FA and 60% GGBFS combination can reach the minimum fmʹ of 10.34 MPa (1500 psi) at 28 days.

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2. The limit on masonry grouts to have a maximum pozzolan content of 40% by mass of the binary cement and a total content of pozzolan and GGBFS of less than 70% by mass of the ternary cement should be increased. New limits should be 65% by mass of the binary cement and 85% by mass of the ternary cement. 3. The strength evolution of the prisms with grouts containing high volumes of SCMs is not as pronounced as the strength evolution of the grouts themselves. 4. A lower estimate of the ultimate strength of grouted prisms constructed with grouts containing high volumes of SCMs can be obtained by multiplying the strength measured at 14 days by 1.2 and 1.3 for prisms with binary and ternary grouts, respectively. AUTHOR BIOS

ACI member Fernando S. Fonseca is an Associate Professor at Brigham Young University (BYU), Provo, UT. He received his BS and MS in civil engineering from BYU in 1987 and 1988, respectively, and his PhD from the University of Illinois at Urbana-Champaign, Champaign, IL, in 1997. His research interests include masonry structures and materials. Scott M. Watterson is a Project Engineer for the structural engineering firm, Cefali & Associates, specializing in earth retention systems. He received his BS from California State University, Northridge, Los Angeles, CA, in 2010, and his MS from BYU in 2011. Kurt Siggard is the Executive Director of the Concrete Masonry Association of California and Nevada, Citrus Heights, CA. He is a graduate of BYU. His research interests include masonry materials, safe and sustainable structures, and furthering the research of concrete masonry products.

ACKNOWLEDGMENTS

This research would not have been possible without the support of the various donors: P. Jahn from Ash Grove Packaging; R. Child from Child Enterprises; K. Hatfield from Doyle Hatfield Masonry; J. Johnson and C. Bedford from Headwaters; H. Holdaway from IMS Masonry; R. Shogren and T. Sherman from Lafarge; W. Ficklin, P. Kamnikar, and G. Travis from Oldcastle; S. Hanks and T. Clark from QUIKRETE; and B. Overson from the Utah Masonry Council. Much appreciation is due to the National Concrete Masonry Association and D. Alan Firmage for providing scholarship monies. BYU graduate students J. Ballard, J. Manuel Salguero Mendizabel, B. Somers, and T. Uyema, undergraduates R. Godfrey and Z. Guo, as well as BYU staff D. Anderson, R. Mayo, and D. Wilson are acknowledged for their substantial assistance with specimen casting and testing.

REFERENCES

1. “Building Code Requirements and Specification for Masonry Structures,” (TMS 402-13/ACI 530-3/ASCE 5-113 and TMS 602-13/ACI 530.1-3/ASCE 6-113), The Masonry Society, Boulder, CO, 2013, 338 pp. 2. Drysdale, R. G., and Hamid, A. A., “Behavior of Concrete Block Masonry under Axial Compression,” ACI Journal Proceedings, V. 76, No. 6, June 1979, pp. 707-721. 3. Cheema, T. S., and Klingner, R. E., “Compressive Strength of Concrete Masonry Prisms,” ACI Journal Proceedings, V. 83, No. 1, Jan.-Feb. 1986, pp. 88-97. 4. Drysdale, R. G., and Hamid, A. A., “Capacity of Concrete Block Masonry Prisms under Eccentric Compressive Loading,” ACI Journal Proceedings, V. 80, No. 2, Mar.-Apr. 1983, pp. 102-108. 5. Steadman, M.; Drysdale, R. G.; and Khattab, M. M., “Influence of Block Geometry and Grout Type on Compressive Strength of Block Masonry,” 7th Canadian Masonry Symposium, Hamilton, ON, Canada, June 1995, pp. 1116-1127. 6. Copeland, R. E., and Timms, A. G., “Effect of Mortar Strength and Strength of unit on the Strength of Concrete Masonry Walls,” ACI Journal Proceedings, V. 28, No. 4, Apr. 1932, pp. 551-562. 7. Shrive, N. G., “The Failure Mechanism of Face-Shell Bedded (Ungrouted and Unreinforced) Masonry,” International Journal of Masonry Construction, V.3, No. 2, 1982, pp. 115-128 8. Page, A. W., and Shrive, N. G., “Concentrated Loads on Hollow Concrete Masonry,” ACI Structural Journal, V. 87, No. 4, July-Aug. 1990, pp. 436-444.

ACI Materials Journal/March-April 2016

9. Romagna, R. H., and Roman, H. R., “Compressive Strength of Grouted and Ungrouted Concrete Block Masonry,” Proceedings of the British Masonry Society, V. 9, 2002, pp. 399-404. 10. Hanle, L. J.; Jayaraman, K. R.; and Smith, J. S., “CO2 Emissions Profile of the U.S. Cement Industry,” 13th International Emission Inventory Conference, Clearwater, FL, June 8-10, 2004, pp 1-14. 11. Mindess, S.; Young, J. F.; and Darwin, D., Concrete, second edition, Pearson Education, Inc., Upper Saddle River, NJ, 2003, 644 pp. 12. Mehta, P. K., and Monteiro, P. J. M., Concrete: Microstructure, Properties, and Materials, third edition, The McGraw-Hill Companies, Inc., New York, 2006, 659 pp. 13. Hogan, F.; Meusel, J.; and Spellman, L., “Breathing Easier with Blast Furnace Slag,” Rock Products: Cement Americas, July-Aug. 2001, pp. 11-15. 14. Malhotra, V. M., and Mehta, P. K., Pozzolanic and Cementitious Materials, V. 1. Taylor & Francis, London, UK, 1996, 191 pp. 15. Davis, R. E.; Carlson, R. W.; Kelly, J. W.; and Davis, H. E., “Properties of Cements and Concretes Containing Fly Ash,” ACI Journal Proceedings, V. 33, No. 5, May 1937, pp. 577-612. 16. ACI Committee 232, “Report on High-Volume Fly Ash Concrete for Structural Applications (ACI 232.3R-14),” American Concrete Institute, Farmington Hills, MI, 2014, 19 pp. 17. Demirdag, S.; Ugur, I.; and Sarac, S., “The Effects of Cement/Fly Ash Ratios on the Volcanic Slag Aggregate Lightweight Concrete Masonry Units,” Construction and Building Materials, V. 22, No. 8, 2008, pp. 17301735. doi: 10.1016/j.conbuildmat.2007.05.011 18. Fonseca, F. S.; Godfrey, R. C.; and Siggard, K., “Compressive Strength of Masonry Drout Containing High Amounts of Class F Fly Ash and Ground Granulated Blast Furnace Slag,” Construction and Building Materials, V. 94, 2015, pp. 719-727. doi: 10.1016/j.conbuildmat.2015.07.115 19. ASTM C476-09, “Standard Specification for Grout for Masonry,” ASTM International, West Conshohocken, PA, 2010, 3 pp. 20. ASTM C90-09, “Standard Specification for Loadbearing Concrete Masonry Units,” ASTM International, West Conshohocken, PA, 2010, 4 pp. 21. ASTM C140-09, “Standard Test Methods for Sampling and Testing Concrete Masonry Units and Related Units,” ASTM International, West Conshohocken, PA, 2010, 16 pp. 22. ASTM C150/C150M-12, “Standard Specification for Portland Cement,” ASTM International, West Conshohocken, PA, 2012, 9 pp. 23. ASTM C618-08, “Standard Specification for Coal Fly Ash and Raw or Calcined Natural Pozzolan for Use in Concrete,” ASTM International, West Conshohocken, PA, 2008, 3 pp. 24. ASTM C989/C989M-14, “Standard Specification for Slag Cement for Use in Concrete and Mortars,” ASTM International, West Conshohocken, PA, 2014, 8 pp.

ACI Materials Journal/March-April 2016

25. ASTM C404-11, “Standard Specification for Aggregates for Masonry Grout,” ASTM International, West Conshohocken, PA, 2011, 3 pp. 26. ASTM C595/C595M, “Standard Specification for Blended Hydraulic Cements,” ASTM International, West Conshohocken, PA, 2010, 13 pp. 27. ASTM C1552-09a, “Standard Practice for Capping Concrete Masonry Units, Related Units and Masonry Prisms for Compression Testing,” ASTM International, West Conshohocken, PA, 2010, 4 pp. 28. ASTM C109/C109M-08, “Standard Test Method for Compressive Strength of Hydraulic Cement Mortars (Using 2-in. or [50-mm] Cube Specimens),” ASTM International, West Conshohocken, PA, 2010, 9 pp. 29. ASTM C1437-07, “Standard Test Method for Flow of Hydraulic Cement Mortar,” ASTM International, West Conshohocken, PA, 2010, 2 pp. 30. ASTM C143/C143M-10, “Standard Test Method for Slump of Hydraulic-Cement Concrete,” ASTM International, West Conshohocken, PA, 2010, 4 pp. 31. ASTM C1019-09, “Standard Test Method for Sampling and Testing Grout,” ASTM International, West Conshohocken, PA, 2010, 5 pp. 32. ASTM C617/C617M-11, “Standard Practice for Capping Cylindrical Concrete Specimens,” ASTM International, West Conshohocken, PA, 2011, 6 pp. 33. ASTM C1314-09 “Standard Test Method for Compressive Strength of Masonry Prisms,” ASTM International, West Conshohocken, PA, 2010, 10 pp. 34. Brown, R. H., and Whitlock, A. R., “Compressive Strength of Grouted Hollow Brick Prisms,” ASTM STP 778, Masonry, Materials, Properties, and Performance: A Symposium, J. G. Borchelt, ed., ASTM International, West Conshohocken, PA, 1982, pp. 99-117. 35. ACI Committee 209, “Guide for Modeling and Calculating Shrinkage and Creep in Hardened Concrete (ACI 209.2R-08),” American Concrete Institute, Farmington Hills, MI, 2008, 44 pp. 36. Zeng, Q.; Li, K.; Fen-Chong, T.; and Dangla, P., “Determination of Cement Hydration and Pozzolanic Reaction Extents for Fly-Ash Cement Pastes,” Construction and Building Materials, V. 27, No. 1, 2012, pp. 560-569. doi: 10.1016/j.conbuildmat.2011.07.007 37. Zhang, Y. M.; Sun, W.; and Yan, H. D., “Hydration of High-Volume Fly Ash Cement Pastes,” Cement and Concrete Composites, V. 22, No. 6, 2000, pp. 445-452. doi: 10.1016/S0958-9465(00)00044-5 38. Song, S.; Sohn, D.; Jennings, H. M.; and Mason, T. O., “Hydration of Alkali-Activated Ground Granulated Blast Furnace Slag,” Journal of Materials Science, V. 35, No. 1, 2000, pp. 249-257. doi: 10.1023/A:1004742027117 39. Drysdale, R. G., and Hamid, A. A., Masonry Structures: Behavior and Design, The Masonry Society, Boulder, CO, 2008, 150 pp.

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ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M19

Compatible Datum Temperature and Activation Energy for Concrete Maturity by Chang Hoon Lee and Kenneth C. Hover Maturity methods are used to predict strength or other mechanical properties of a concrete mixture for a given moisture availability as a function of time and temperature. Temperature sensitivity of a mixture is characterized by datum temperature for the Nurse-Saul (NS) method, and by activation energy for the Freiesleben-Hansen and Pederson method (FHP). While these methods and their defining parameters were independently developed, the parameters are nevertheless interdependent, as a change in a concrete mixture that affects temperature sensitivity as expressed by datum temperature will also be reflected in a change of activation energy, and vice versa. This paper addresses the temperature- and mixturedependent relationship between datum temperature and activation energy, using both a closed-form equation based on chemical kinetics and experimental data where values of datum temperature and activation energy were found to provide the best fit to strength data. Best-fit results validate the closed-form solution. Keywords: activation energy; concrete strength; datum temperature; mathematical modeling; maturity; physical chemistry; temperature effects.

INTRODUCTION The maturity method is used to predict concrete properties based on time-temperature history. The widely-implemented methods adopted by ASTM C10741 are based on the work of Freiesleben-Hansen and Pederson2 (referred to herein as FHP), and by Nurse3 as modified by Saul4 (referred to herein as NS). The NS method assumes that the rate of the chemical and physical processes that lead to development of concrete properties (most often, strength) increases linearly with concrete temperature, and this temperature sensitivity is characterized by datum temperature Td, interpreted as the temperature below which concrete does not gain strength. Approximately 30 years after the introduction of NS, the FHP method applied a more fundamental approach based on the work of Arrhenius,5 preceded by Maxwell and Boltzmann,6 that assumed a nonlinear relationship between concrete temperature and the rate of development of concrete properties, and characterized temperature sensitivity by activation energy Ea. As documented herein and elsewhere,7 whether the linear NS or the nonlinear FHP maturity method more accurately predicts concrete strength depends on the specific concrete mixture and time-temperature record in question, and in particular, on the form of the relationship between the rate constant and concrete temperature. The more linear the relationship between rate constant k and concrete temperature T over the range of expected concrete temperatures, the more applicable are the underlying assumptions of the NS method. As the fundamental relationship between k and T tends towards exponential, the more applicable are the ACI Materials Journal/March-April 2016

assumptions behind the FHP method as long as the range of concrete temperatures remains within the range over which the exponential k-T function remains valid. As the range of expected concrete temperatures becomes narrower, especially at colder temperatures, the differences between the two methods diminish. Users should therefore compare methods to find that which produces the most useful results for their conditions. (Note that various proprietary systems are often pre-programmed for only one of the two methods.) In conducting such comparisons, or in exploring a change from one method to the other, it is essential that Td (in NS method) and Ea (FHP) equivalently represent the inherent temperature sensitivity of strength development for the specific concrete mixture over the time period and temperature range in question. (In either case, the fundamental temperature sensitivity itself is a function of the chemistry and proportions of the particular blend of cementitious materials, and therefore independent of any maturity or computational method.) This paper describes how to find compatible values of Ea and Td that equivalently characterize this inherent temperature sensitivity. Following a brief review of fundamental concepts, the authors derive closed-form equations that allow conversion from any value of Ea used in the nonlinear FHP model to a compatible value of Td that when used in a linear NS model will predict the same relative rate constant at the same temperature. This generic development is based entirely on chemical kinetics as described by basic rate laws and the rate constant, k, and is not only applicable beyond values typically encountered in cement and concrete, but to a wide range of temperature-dependent rate growth models in general. As previously published,7 the authors applied a numerical approach to experimental data from four different studies to find values of Ea and Td that minimized prediction error for compressive strength by both the FHP and NS methods. This current paper merges the analytical and experimental findings. RESEARCH SIGNIFICANCE Mixture and time-temperature history dictate whether NS or FHP maturity methods produce most accurate predictions.7 Users should compare methods for best results. In such comparisons both methods must equivalently characterize the inherent temperature sensitivity of strength ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-161.R1, doi: 10.14359/51688639, received June 2, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

197

development for the specific concrete mixture. For a given mixture for which a value of Ea has been found to effectively describe temperature sensitivity, a limited range of compatible values of Td over a limited range of concrete temperature Tconc describes temperature sensitivity with similar effectiveness. One cannot independently select values of Ea and Td and expect similar accuracy of prediction from the NS and FHP methods. BACKGROUND Rate constants in concrete maturity method A key to understanding the difference and the fundamental similarity of the NS and FHP methods is recognition that for any chemical reaction, the effect of temperature on reaction rate can be characterized by a coefficient known as the rate constant, k.8 For example, if reactants A and B combine to form products C and D in accordance with aA + bB → cC + dD, the rate at which the concentration of C changes with time is given by8

d [C ] m n = k [ A] [ B ] (1) dt

where reactants and products are expressed as their concentrations in units such as mol/L, d[C]/dt is the rate of change of concentration of product C in units such as mol/L/h, and k is the rate constant in units of Lm+n– 1/mol/h. (Units in moles can of course be converted to mass units.) Also, m + n is the total order of a given reaction. By monitoring consumption of reactants and formation of products of this chemical process over time and at various temperatures, one determines the relationship between k and temperature. The rate constant is often written as k(T) to denote temperature dependency. Applying this concept to one form of cement hydration, where 2C3S + 11H2O → C3S2H8 + 3Ca(OH)2,9 one could monitor production of C3S2H8 (CSH) over time at various temperatures to find k for cement hydration and its dependency on temperature. Maturity methods in concrete apply this same concept with a subtle yet important difference. The output of temperature-dependent reactions of cementitious materials is not measured in terms of mass or concentration of hydration products, but in terms of concrete properties that result from those chemical reactions, as influenced by aggregates and other unique characteristics of a given mixture. Thus, the “hybrid” or “apparent” rate-constant approach in the concrete maturity method essentially assumes that the reactants are cementitious materials and water, and the product is compressive strength (or flexural strength, elastic modulus, setting behavior via penetration resistance, among others) While there are multiple methods for experimentally determining rate constant from strength development in mortar and concrete,6-8 ASTM C1074 Annex A11 suggests a hyperbolic curve-fit as a means for extracting rate constant k from compressive strength time data collected over a series of constant concrete temperatures. Pinto and Hover10 used this same approach to extract rate constants from penetration resistance measurements of setting behavior. In all such

198

Fig. 1—Rate constants as function of temperature for Carino’s mortar strength versus time data,11 showing linear and nonlinear curve fits. Linear fits are partitioned over various ranges of temperature. (Note: °F = [1.8 × °C] + 32.) cases, the term “apparent rate constant” would be more technically correct to differentiate it from that obtained directly from mass or concentration of products in conventional chemical kinetics, but following common usage in concrete technology, the adjective “apparent” is not used herein. Quantifying influence of temperature on absolute and relative rate constant Following Carino’s clear and complete example,11 the five points plotted in Fig. 1 are (T, k) data pairs obtained from mortar strength versus time data per ASTM C1074 Annex A1.1 One can quantify the influence of temperature on rate constant (related to the rate of strength gain) by fitting these data in a variety of functional forms, ranging from straight lines to more complicated nonlinear power or exponential models. Herein is the fundamental difference between NS (linear) and FHP (nonlinear) maturity methods. In Fig. 1, the smooth, continuous curve is fitted by the Arrhenius equation,5,12 and is the exponential model employed in the FHP method. (For the specific nonlinear curve shown, Ea = 42.2 kJ/mol, and the pre-exponential factor, A = 9.9 × 106, the significance of which will be discussed later.) The straight lines are the best linear fits over selected temperature ranges, as assumed in the NS method. In algebraic terms, the first fundamental difference between the NS and the FHP methods is that the NS method implicitly assumes a linear relationship between rate constant, k(T) and (T – Td) as shown in Eq. (2), while the FHP method is based on the Arrhenius equation2,5,11 shown in Eq. (3)

k(T) = β(T – Td)

(2)

 E  k (T ) = A exp  − a  (3)  RT  ACI Materials Journal/March-April 2016

Table 1—Results of linear and nonlinear regressions for data shown in Fig. 1 Regression method (range of temperature)

A, 1/day

Arrhenius (5.5 to 43°C) 9.92 × 106

Ea, kJ/mol

Td, °C

42.2

R2

SE, 1/day

0.998

0.042

Linear (5.5 to 43°C)

4.5

0.922

0.121

Linear (5.5 to 23°C)

–3.9

0.995

0.011

Linear (23 to 43°C)

14.9

0.991

0.047

Note: °F = (1.8 × °C) + 32.

where k(T) is the rate constant; β is the constant slope of the NS line; Td is the NS datum temperature; A is what Arrhenius called the pre-exponential factor; Ea is activation energy; R is the ideal gas constant; and T is concrete temperature. (Temperature in Eq. (3) must be in degrees Kelvin, but in Eq. (2), as only a temperature difference is used, values of T can be in K, C, or F and is accommodated in the value of β.) Ea is interpreted as the required level of kinetic energy possessed by the reactants in order for the reaction to occur. A is interpreted as the number of potential molecular “collisions” per unit time, of which only a fraction will result in products. Arrhenius derived his equation based on the Van’t Hoff equation and the Maxwell-Boltzmann distribution of energy for ideal gas molecules at a given temperature.12 Figure 1 shows another key distinction between NS and FHP: the Arrhenius curve asymptotically approaches a rate constant of zero at infinitely low temperatures. The linear models all reach zero rate constant at a discrete temperature, below which no strength gain is assumed to occur. These x-intercepts for the linear models correspond to what Saul4 called the “minimum temperature,” later termed the datum temperature by Plowman.13 (Although Nurse3 did not deliberately employ the notion of a datum temperature, he developed his method for high-temperature steam-curing and implicitly made 0°C [32°F] the baseline or datum value.) Table 1 shows results from fitting the data in Fig. 1, obtained per the methodology in References 1, 11, and 14. For this sample data, over the full range of temperature, the Arrhenius (FHP) model provides a better fit than the linear NS model (R2 = 0.998 versus R2 = 0.922), as is often but not always the case. However, as the range of temperature of interest becomes narrower, the error diminishes as the linear model more closely approximates the exponential model (the lowest SE [standard error] is obtained with the linear model over the range 5.5 to 23°C [42 to 73°F]). The capacity of the Arrhenius equation to provide a better fit to nonlinear rate-constant data has often been considered an advantage of the FHP maturity methods.11,15-18 Nevertheless, the NS variation of the maturity method has been widely and successfully used, noted by its frequent use in maturity meters.19,20 Generally speaking, the FHP method would be better suited to rate-constant data more accurately represented by a nonlinear model. NS could yield acceptable, or in some cases more accurate, predictions for rate constants more linearly related to temperature over the range of interest. Practical implementation of the relationship shown in Fig. 1 is challenging, however, because calculation of absolute k at any temperature Ti requires values for both Ea and A ACI Materials Journal/March-April 2016

Fig. 2—Series of Arrhenius curves for Ea = 42.2 kJ/mol with Variable A. (Note: °F = [1.8 × °C] + 32.) in Eq. (3), and values for both Td and β in Eq. (2). Figure 2 illustrates this for the data of Fig. 1. Three curves are plotted, each with the same Ea but for three different values of A, selected from within the 95% confidence interval obtained via curve-fitting the initial rate constant versus temperature data. As seen in this example, the value of A obtained via nonlinear regression is statistically imprecise, with multiple orders of magnitude separating the lower- and upper-bound confidence interval at the 95% confidence level—that is, 2.6 × 106 to 3.8 × 108 for the example of Fig. 1. As is seen in the figure, the variability of A makes it difficult to compute an absolute rate constant with confidence. A is nevertheless calculable from the same dataset used to produce Ea per the ASTM C1074 Annex A1 analysis, but extracting A from rate constant data is not currently part of standard practice as described by ASTM C1074. Likewise, the NS method as currently implemented does not include capture of the value of β, which is the slope required to compute k from a linear model for which Td is known. The best-fit value of Td, however, is the x-intercept at which k = 0 remains valid, independent of the value of β. Given this general inability to reliably compute absolute rate constants, standard practice has implicitly and explicitly adopted a relative rate constant approach. This is clear in the FHP method, where the incremental equivalent age over a given time interval, Δt, is computed by multiplying the time interval by the age conversion factor, calculated precisely as the rate constant at Ti divided by the rate constant at Tref. The relative rate constant is therefore another and more generic name for the ASTM C1074 age conversion factor.11 As seen by the definition of these rate constants in Eq. (3), this relative approach neatly solves one problem by the fact that the pre-exponential factor A cancels out of the result. Thus in the FHP method, the cumulative equivalent age at any time is expressed as relative to the actual clock or calendar age for concrete continuously maintained at the reference temperature, and predictions are therefore based on reaction rates at Ti in linear proportion to measurements observed at Tref. 199

The “relative” nature of the NS method is subtly embedded in the absence of the slope term β in the conventional NS maturity calculation as in ASTM C1074 n

M (T , t ) = ∑ (Ti − Td )∆t (4)



i =1

The value inside the summation in Eq. (4) is equivalent to the fundamental kinetics expression of Eq. (2) only when β = 1. (This is not to imply that Nurse or Saul necessarily intended their model to represent the pure kinetics approach.) For most practical data sets, this not only significantly overestimates β and the resulting rate constant, but it makes the implied value of β the same for all concrete mixtures, when in fact β is as unique and mixture-dependent as is Td. However, within any given concrete, assuming β = 1 will proportionately increase the computed rate constants for all data sets from the same concrete, including the rate constant at the reference temperature, still yielding appropriate values for relative rate constants. Thus when using either the NS or FHP methods, concrete strength or other properties are predicted on the basis of comparison with concrete maintained at a reference temperature. Because relative rate constants are therefore fundamental to current maturity practice, attention is now given to finding values of Ea and Td that will each return the same value of relative rate constant for a given concrete at a given temperature. DERIVATION OF COMPATIBILITY OF DATUM TEMPERATURE AND ACTIVATION ENERGY Relative rate constant (age conversion factor)based compatibility between Td and Ea As pointed out earlier, relative rate constants are explicitly included in the FHP method, and implied in the conventional NS method. However, De Schutter21 reported Rastrup’s proposed modification of NS methods to fully embrace the notion of relative rate constant, termed the “Affinity Ratio,” γ(T) as shown in Eq. (5) as evolved from Eq. (2) for rate constant at T relative to that at Tref.

γ NS (T ) =

β (T − Td )

(

β Tref − Td

)

=

T − Td (5) Tref − Td

Rastrup’s affinity ratio has been subscripted herein to identify it with the NS method, and to distinguish it from the FHP age conversion factor γFHP(T) shown in Eq. (6)  E 1 A exp  − Ea / ( RT )  1   γ FHP (T ) = k (T ) = = expp  − a  − 

( )

k Tref

(

)

A exp  − Ea / RTref 



R T

Tref  

(6) Equation (6) is also shown as derived from Eq. (3), with the final result as found in ASTM C1074. Examples of these relationships are shown in Fig. 3, which contains a feature unique to the relative rate constant development: both Rastrup’s and FHP relative rate constants (affinity ratios or age conversion factors) must pass through the fixed point Tref, 1. A relationship can be derived for the compatibility 200

Fig. 3—Concepts of equal age conversion factor for four different concrete temperatures with Tref = 20°C (68°F). (Note: °F = [1.8 × °C] + 32.) of values of Td and Ea for which the relative rate constant used in the NS-Rastrup method will be identical to the relative rate constant obtained via the FHP method at any concrete temperature; that is, values of Td and Ea that will make γNS(T) = γFHP(T).

 E T − Td = exp  − a Tref − Td  R

1 1 T − T  ref

   (7)  

which upon rearrangement of terms leads to

Tdc (Ti ) =

 E where m = exp  − a  R

Ti − mTref 1− m

(8)

1 1   T − T  .  i ref   

Similarly, Eq. (8) can be inverted to derive compatible activation energy from known Td

 Ti ⋅ Tref Eac = R ⋅   Ti − Tref

  Ti − Td   ⋅ ln  T − T  (9)   ref d 

In Eq. (8) and (9), the variable names Tdc and Eac denote output of the functions for which the user inserts available values of either Ea or Td, and obtains compatible parameters that will produce equivalent relative rate constants at the specified temperature. Equation (8) is plotted in Fig. 4 and results tabulated in Table 2 for the range of concrete temperature from 0 to 40°C (32 to 104°F) and for seven discrete values of activation energy Ea. Note first that for concrete temperature in the range of 10 to 13°C (50 to 55°F), a value of Td = 0°C (32°F) is compatible with values of 40 ≤ Ea ≤ 45 kJ/mol. This means that at these values, nearly identical age conversion factors are computed for either the NS or FHP maturity methods. Note also that these are the default values of Td and Ea recommended by ASTM C1074. ACI Materials Journal/March-April 2016

Table 2—Tabulated compatible datum temperature Tdc as function of concrete temperature Ti and of various values of Ea by Eq. (8) with Tref = 20°C (68°F) Ea, kJ/mol

*

Ti, °C

30

35

40

45

50

55

60

0

–13.7

–10.72

–8.58

–6.97

–5.72

–4.73

–3.94

5

–10.9

–7.81

–5.52

–3.77

–2.4

–1.3

–0.4

10

–8.3

–5.12

–2.72

–0.86

0.61

1.81

2.79

15

–6.0

–2.65

–0.16

1.78

3.32

4.58

5.63

20

*

–3.8

*

–0.4

2.2

4.1

5.7

7.0

8.1*

25

–1.8

1.7

4.24

6.24

7.84

9.14

10.23

30

0.1

3.5

6.1

8.09

9.68

10.96

12.03

35

1.8

5.2

7.75

9.72

11.27

12.51

13.53

40

3.3

6.7

9.22

11.13

12.62

13.81

14.77

*

*

*

*

*

Computed by Eq. (10) at reference temperature.

Note: °F = (1.8 × °C) + 32.



Fig. 4—Temperature-dependent datum temperature expressed in Eq. (8). (Note: °F = [1.8 × °C] + 32.) However, at concrete temperatures beyond the range shown, the most appropriate pairs of Td and Ea diverge from ASTM recommendations. Further, at colder concrete temperatures (–6 to –4°C [21 to 25°F]), essentially identical age conversion factors are obtained via Saul’s datum temperature of –10.5°C (13°F) and C1074’s recommended range of activation energy. It is noted that Eq. (8) is undefined when concrete temperature is equal to the reference temperature, as indicated by the open circles in Fig. 4. This problem can be solved two ways. First, by incrementally approaching Tref from above or below the desired value can be approximated to any desired degree of precision. Second, by recognizing that at Tref, the linear function must be tangent to the nonlinear Arrhenius curve, one can derive the expressions given in Eq. (10) and (11) for compatibility at that one temperature.

( )

Tdc Tref = Tref −

Tref 2 Ea / R

(10)

ACI Materials Journal/March-April 2016

( )

Eac Tref =

RTref 2 Tref − Td

(11)

EXPERIMENTAL VALIDATION OF DERIVED RELATIONSHIPS As reported elsewhere,7 the authors analyzed four separate data sets from the literature and compiled in their own laboratory. Mixtures ranged from 307 to 522 kg/m3 (519 to 882 lb/yd3) cement and concrete temperatures ranged from –3 to 49°C (37 to 120°F). For each data set, a time-temperature history was available along with compressive strength data at various ages. The authors compared actual strength to that predicted by both NS and FHP methods, and in each case systematically varied Td and Ea over a wide range of values to minimize the difference between actual and predicted strength. For strength at any one age (3 days, for example), it was possible to find unique values of Td and Ea that returned zero difference between measured and predicted strength. In cases of attempting to minimize these differences over multiple ages (1, 3, 7, and 28, days, for example), the objective was to find values of Td and Ea that minimized the standard error for all ages combined. Zero difference was generally not possible for multiple ages due to the fact that a given set Td and Ea that worked effectively at one age was not likely to represent the same concrete as effectively at another age at test. In contrast to the kinetic, first-principles approach taken in the derivations herein, in minimizing the error in strength prediction, values of Td and Ea were iteratively varied as regression coefficients until the objectives (zero relative error or minimum standard error) were achieved. This process ignored both the physical chemistry interpretations of Ea and Td and the recommended default values in ASTM C1074, yet nevertheless returned matching pairs of “minimum error” values of Ea and Td reported herein in Table 3. These values of Td and Ea are therefore unbiased estimators. In contrast, the proposed compatibility relationships in this paper are based on and constrained by underlying physical chemistry and the mathematical formulations of rate constants and age conversion factors. Reference 7 201

Table 3—Td and Ea producing zero error at 3 days or minimum standard error (SE) for all test ages7 3-day strength Data label

Td, °C

Overall strength

Ea, kJ/mol

Td, °C

SE, MPa KC

Annex A1

Ea, kJ/mol

SE, MPa

Tconc range

Td, °C

Ea, kJ/mol

22

KC1 (–3°C)

–4.9

54.0

–4.6

0.33

71.1

1.80

L

–10.3

65.6

KC2 (–5°C)

2.6

63.3

–1.7

3.29

47.3

2.76

U

1.9

38.4

KC3 (13°C)

5.1

50.1

–1.9

2.82

35.0

2.70

A

–4.5

47.8

KH

22

KH2 (32°C)

–7.8

20.5

–20.8

1.39

15.4

1.37

L

–109.2

30.7

KH3 (41°C)

–21.5

14.2

–70.8

3.79

7.7

3.79

U

5.9

6.1

KH4 (49°C)

–202.2

3.3

<–273 (354.4)

<6.06 (5.93)

<0 (–2.8)

<5.96 (5.92)

A

–13.6

18.1

–27.6

16.6

LZ

17

LZ1 (–2°C)

–6.5

45.1

–11.5

5.91

35.9

1.28

L

LZ2 (14°C)

<–273 (73.3)

<0 (–12.7)

<–273 (98.6)

<2.82 (2.76)

<0 (–8.7)

<5.96 (5.92)

U

6.9

38.5

LZ4 (43°C)

–8.4

19.3

–56.4

1.28

8.9

1.28

A

–4.6

27.6

LZv (15 to 47°C)

–7.1

21.6

–10.3

1.51

19.3

1.52

—

—

—

LH

7

LH1 (–1°C)

–2.4

78.9

–3.6

2.18

64.7

2.19

L

–5.6

55.2

LH2 (11°C)

–1.3

38.4

–3.1

2.19

36.4

2.13

U

0.2

29.5

LH4 (42°C)

<–273 (245.8)

<–273 (68.5)

<3.65 (2.57)

<0 (–21.9)

<3.47 (2.58)

A

–2.2

43.2

<0 (–3.7)

Notes: Ea = 33.2 kJ/mol per Eurocode 2, fib Model Code ; Ea = 40 to 45 kJ/mol per ASTM C1074 Appendix X1, Type I cement without additive; Td = 0°C (32°F) recommended per ASTM C1074 Appendix X1, Type I cement without additive, 0 to 40°C (32 to 104°F) curing temperature; Td = –10.5°C (13°F)4; L indicates values computed by ASTM C1074 Annex A1 with three coolest subsets; U indicates values computed by ASTM C1074 Annex A1 with three warmest subsets; A indicates values computed by ASTM C1074 Annex A1 with all four subsets; KC and KH were collected from Reference 25; LZ17 and LH7 data sets were experimentally obtained in the authors’ laboratory; values in parentheses in first column indicate average concrete temperatures; °F = (1.8 × °C) + 32. 23

24

includes a fuller comparison of experimentally obtained values of Td and Ea with the ASTM C1074 default recommendations and the results of analysis per ASTM C1074 Annex A1. A comparison of experimental results the analysis performed in this paper follows. Figure 5(a) shows the surface created by Eq. (8), with superimposed values of Td and Ea that had been experimentally found to result in zero relative error at 3 days for four specific data sets. Similarly, Fig. 5(b) shows the same surface of theoretical compatibility between datum temperatures with experimental values of Td and Ea that minimize prediction errors combined over test ages varying from 1 to 28 days. Table 4 summarizes the deviation statistics between compatible datum temperatures computed by Eq. (8) and independently estimated Td that minimized prediction error for both 3-day strength and at all test ages combined. In general, the difference between values experimentally determined, best-fit values of Td, and those computed by Eq. (8) ranged from 0.1 to 3.5°C (32.2 to 38.3°F). (In each case, computed Td were based on best-fit Ea.) An exception is an apparent outlier in the KH data sets shown in Table 3, to be discussed. In regard to apparent anomalies, however, recall the wellknown effect of high-temperature curing that generally reduces later-age concrete strength. From a pure chemical kinetics perspective, this is interpreted as a decreased rate constant for strength gain associated with an increase in concrete temperature, and will be manifested in a typical Arrhenius plot as a negative Ea. This same effect is modeled 202

in the NS method by a datum temperature that is greater than concrete temperature, registering as a negative increment of maturity for an increment of time at such high temperature. Given the theoretical admissibility of such non-traditional values of Ea and Td, it is interesting to a) keep such values in the Table 3 data set; and b) recognize that Eq. (8) is actually a three-dimensional hyperbola, with an upper branch fully shown in Fig. 6. The apparently anomalous data points are well fitted by the upper branches of the hyperbolic surfaces of Eq. (8), with associated deviations shown in Table 4. DISCUSSION In both the relationships defined by Eq. (8) and (9), compatible values of Td and Ea depend on the concrete temperature, leading to recognition that Td and Ea are themselves temperature-dependent. The concept of temperature-dependent activation energy is most fundamentally observed in a nonlinear Arrhenius plot, but was observed in specific application to concrete maturity by Freiesleben-Hansen and Pederson,2 followed by other researchers.17 Figure 7 shows the FHP proposed temperature-dependent Ea, drawn against a background of solutions to Eq. (9) for compatible values of Ea at given values of Td. The plot includes the 0°C (32°F) C1074 default, Saul’s proposed –10.5°C (13°F), and Snyder and Bentz’s26 observation that hydration continues down to –30°C (–22°F). Given the operational assumption that Td represents the concrete temperature at which reaction rate is equal to zero, it is of interest to compute values of Ea that would be compatible with these three proposed values. ACI Materials Journal/March-April 2016

Fig. 5—Comparison of compatible Td with concrete temperature and Ea with Td at minimum prediction error for Ea ≥ 0 and Tconc ≥ Td. (Note: °F = [1.8 × °C] + 32.) Table 4—Standard error between compatible Td computed by Eq. (8) and optimum Td found by analysis of experimental data Analysis

KC

KH

LZ

LH

Average

3 days: Ea ≥ 0 and Tconc ≥ Td

1.83

1.69

3.22

0.57

2.14

All test ages: Ea ≥ 0 and Tconc ≥ Td

0.55

1.56

3.52

0.12

2.07

3 days: No range constraint

1.83

1.69

2.85

1.82

2.17

All test ages: No range constraint

0.54

20.12

3.11

0.35

9.82

Freiesleben-Hansen and Pederson’s2 proposed temperature-dependent Ea is a reasonable bilinear representation of values obtained with the ASTM-suggested 0°C (32°F) datum (thus demonstrating a common fundamental basis for the FHP and NS methods). As reported earlier, this same ASTM datum temperature supports an ASTM suggested value of Ea of 40 to 45 kJ/mol in the temperature range of approximately 10 to 18°C (50 to 64°F). At a concrete temperature of approximately 23 to 25°C (73 to 77°F), data in the figure support the commonly used Eurocode23 and FIB24 value of 33 kJ/mol, with compatible Ea dropping to less than 25 kJ/mol as concrete temperatures approach 40°C (104°F). If a value of Td of –10.5°C (13°F) were known to be an effective index to the temperature sensitivity of a given mixture, Ea = 33 kJ/mol would only become a compatible value at a concrete temperatures below approximately 2°C (36°F), and Ea = 40 to 45 kJ/mol would return equivalent relative rate constants only at much colder temperatures. Finally, the activation energy associated with a physical datum temperature of –30°C (–22°F) is approximately constant at approximately 15 kJ/mol. It is interesting that about this same value (10 to 15 kJ/mol) has been reported as indicative of diffusion processes in aqueous pore solutions.27 Given this temperature dependency of both Td and Ea, even though maturity calculations commonly assume a fixed value of Ea or Td, regardless of concrete temperature, these results suggest improved accuracy of maturity-based predictions of concrete properties by considering a temperature-dependent Ea or Td. Thus, the more accurately one can bound ACI Materials Journal/March-April 2016

the expected concrete temperature (or temperature range), the more accurately one can select appropriate values of Td and Ea. This is not a significant problem for nearly constant temperature conditions as may pertain in manufacture of some concrete products. Given the typical variability of concrete temperature on most construction sites, however, this can be accommodated by various averaging methods, the simplest of which is to compute the arithmetic average of concrete temperature over the test age of interest, or over various time steps. In this regard, it can be shown that if one were to adjust the value of Td in accordance with Eq. (8), for the average temperature over each time step in a typical NS maturity calculation, the results of the NS method will converge on the results of FHP for a given Ea. This is also true for a convergence of FHP upon NS when Ea is adjusted for each time step in accordance with Eq. (9). In addition to temperature-dependent values of Td and Ea, it is likewise not uncommon to fix those values over the entire duration of the concrete curing (strength gain) period, but these observations likewise suggest that these parameters are time-dependent as well. This is most clearly seen in the difference between best-fit values of Td and Ea for 3 days versus the 1 to 28-day overall results. Previous work7 and the results presented herein not only demonstrate the benefit of tuning NS and FHP maturity methods by adjusting values Td and Ea, but also demonstrate that when these parameters equivalently represent the actual temperature sensitivity of a specific concrete mixture over the temperature range of interest, both methods can provide equivalent accuracy. Key insight to the effectiveness of NS versus FHP is inferred from the relationship between rate constant k and temperature T (example in Fig. 1), which is obtained from strength, time, and temperature data independent of either the NS or FHP methods. When this relationship is well described by a linear function, NS is a fundamentally valid model; and when an exponential function is a better fit to the data, the FHP method is more appropriate. But just as important is the linearity of the k-T data over the particular range of interest, and the narrower that range, the more similar the two methods become due to the ability to approximate a curve with a line as the temperature interval becomes shorter. Similarly, and as seen in Fig. 1, even an exponential k-T relation becomes approximately linear at 203

Fig. 6—Comparison of compatible Td with concrete temperature and activation energy with Td at minimum prediction error with no condition for range of Td and Ea. (Note: °F = [1.8 × °C] + 32.) are not obtained with the alternate method, there is no advantage to changing methods. Finally, the work presented herein is subject to the same limitations as the maturity approach in general.11 In particular, the influence of temperature on the rate of strength gain is contingent on the availability of moisture. Higher temperatures accelerate hydration but diminish later-age strength. Many factors influence strength development and strength testing beyond the kinetics of hydration. Even apparently sophisticated kinetic models (such as the Arrhenius basis for the FHP method) are approximations and their applicability to other than ideal gases and dilute solutions have been questioned.6 The reality of batch-to-batch and day-to-day variability of the concrete mixture can be difficult to accommodate in maturity predictions.

Fig. 7—Temperature-dependent activation energy. (Note: °F = [1.8 × °C] + 32.) cooler temperatures, further reducing the difference between the NS and FHP models. As an example of using the techniques shown here, consider the need to use the NS method (perhaps because it is programmed into a maturity meter) for a mixture containing ground-granulated blast-furnace slag (GGBFS). From the function proposed by Barnett et al.,28 a value of Ea can be obtained (Fig. 8(a)), but the user needs to find a compatible value of Td. Figure 8(b) shows compatible values of Td, computed by Eq. (8) and (10) for average concrete temperatures T = 10, 20, 30, and 40°C (50, 68, 86, and 104°F) with Tref = 20°C (68°F). In other applications, values of activation energy reported in the literature can be converted to the compatible datum temperature (or vice versa) without the need for additional laboratory work per ASTM C1074 Annex A1. The need to adjust parameters or to compare results of NS versus FHP maturity models depends entirely on the user’s logistical capacity to collect and analyze time, temperature, and strength data, and the accuracy required for a given application. If the user is getting acceptable results with either method, and has demonstrated that even better results

204

CONCLUSIONS This research proposed a relationship among concrete temperature, datum temperature for the NS method, and activation energy for the FHP method to achieve equivalent relative rate constants for application to concrete maturity. The derived relationship is validated by comparing experimental and analytical results. The following conclusions were drawn. 1. A valid comparison of NS versus FHP methods requires that values of Td and Ea each fairly represent the temperature sensitivity of the concrete in question. These findings provide guidance to the user in the selection of equivalently effective parameters for the purposes of comparison, when switching from one method to the other, or when transforming a parameter acquired by one method for use in the other method. 2. ASTM C1074 default values for Td (0°C [32°F]) and Ea (40 to 45 kJ/mol) produce equivalent values of relative rate constant within the range of concrete temperature of approximately 10 to 13°C (50 to 55°F), with reduced equivalency at higher and lower temperatures. 3. Although ASTM C1074 suggests the default value of Td = 0°C (32°F) up to Tconc = 40°C (104°F); at this concrete temperature the compatible value of Ea would be approximately 25 kJ/mol rather than the recommended default of 40 to 45 kJ/mol. 4. Seemingly disparate recommendations for Ea can be rectified when viewed as being influenced by concrete ACI Materials Journal/March-April 2016

Fig. 8—(a) Barnet’s activation energy28 with varying GGBFS content; and (b) computed compatible datum temperatures by Eq. (8). (Note: °F = [1.8 × °C] + 32.) temperature. For concrete for which Td = 0°C (32°F) has been shown to lead to acceptable strength predictions, at concrete temperature above 20°C (68°F) (warm and hot weather) a value of 33 kJ/mol (Eurocode2,23 FIB24) is likely to be effective, while 40 to 45 kJ/mol (ASTM1) may be more useful in cool weather with concrete temperature closer to 10°C (50°F). 5. While the paper has focused on the compatibility of values of Td and Ea, such compatibility is not absolute. Figure 4 displays ranges of concrete temperatures over which various values of Td and Ea similarly represent the temperature sensitivity of a concrete mixture. But the degree of similarly needed to consider the final results of the maturity calculations equivalently useful or reliable depends on the uncertainty of all other steps in the process, and the user’s need for precision. In many cases, it is likely that compatible values of Td and Ea that are sufficiently close to the values suggested here will meet the user’s needs. 6. When the criticality of the application warrants, users should perform their own analyses, varying strength predictions by means of systematically varying values of Td or Ea and comparing the results with measurements to find their own optimal parameters.7 In conducting such a process, the values given by Eq. (8) or (9) or associated figures can furnish a useful starting point for the iterative search. 7. Apparent anomalies in best-fit values of Td and Ea can make sense on the basis of pure kinetics. For example, the reduction in later-age strength that often results from high concrete temperature gives rise to best-fit values of Ea that are negative, and best-fit values of Td that are higher than Tconc. In this case, both unusual values appropriately reflect the fact that higher temperatures lead to lower later-age strength—that is, reaction rate is lower at a higher temperature. Within the data sets reported here, apparent anomalies or outliers were well-fitted by the full plot of the surfaces defined by Eq. (8). 8. The relationships of Eq. (8) and (9) essentially allow for translation between linear and Arrhenius-based models for the influence of temperature on a generic growth process. Utility is therefore not constrained to cement and ACI Materials Journal/March-April 2016

concrete related applications. Within the field of cement and concrete, however, the results are applicable to prediction of the development of concrete properties in addition to compressive strength. AUTHOR BIOS

Chang Hoon Lee received his PhD from the School of Civil and Environmental Engineering, Cornell University, Ithaca, NY, and his BS and MS at Korea University, Seoul, South Korea. His research interests include mathematical modeling of property transitions and temperature-time effects of cement-based materials. Kenneth C. Hover, FACI, is Professor and Weiss Presidential Fellow at Cornell University. He is a member of ACI Committees 301, Specifications for Concrete; 305, Hot Weather Concreting; 306, Cold Weather Concreting; and ACI Subcommittee 318-A, General, Concrete, and Construction (Structural Concrete Building Code). He is an ACI Past President.

REFERENCES

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ACI Materials Journal/March-April 2016

ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M20

Performance of Full-Scale Self-Consolidating Rubberized Concrete Beams in Flexure by Mohamed K. Ismail and Assem A. A. Hassan This research investigated the performance of full-scale self-consolidating rubberized concrete (SCRC) and vibrated rubberized concrete (VRC) beams in flexure. The beam mixtures were developed with a maximum possible percentage of crumb rubber (CR) (0 to 50% by volume of sand) while maintaining acceptable fresh properties and minimum strength reduction. The mixture variables included different binder contents, the addition of metakaolin, and the use of air entrainment. The performance of the tested beams was evaluated based on load-deflection response, concrete strain/stiffness, cracking behavior, first crack load, ultimate load, ductility, and toughness. In general, increasing the CR content decreased the mechanical properties, first crack load, stiffness, and self-weight of all SCRC and VRC beams. However, using up to 10% CR enhanced the deformation capacity, ductility, and toughness of tested beams without affecting the flexural capacity. This improvement in the deformation capacity, ductility, and toughness appeared to continue up to 20% CR (but with a slight reduction of the flexural capacity) and then reduced with further increases in the CR content. The results also indicated that although it was possible to produce VRC beams with higher percentages of CR (50% compared to 40% in SCRC), this increased percentage only gave VRC beams an advantage in terms of self-weight reduction, while it had a limited contribution in enhancing the structural performance of the beams. Keywords: beam(s); cracking behavior; crumb rubber; deflection characteristics; flexure capacity; reinforced concrete; self-consolidating concrete.

INTRODUCTION Over the last two decades, waste rubber in concrete has received greater attention due to its availability in large volumes. For example, the applications of waste rubber in concrete in 2011 were estimated to be 1 billion tires produced worldwide.1 The review of literature showed that many studies have been conducted to investigate the performance of concrete with different levels of rubber replacement. Researchers have found that substituting fine and/or coarse aggregates with crumb or shredded rubber particles in concrete enhances its strain capacity (ductility), energy dissipation, damping ratio, impact resistance, and toughness compared to normal concrete using conventional aggregate.1-4 Using rubber can significantly contribute to the development of semi-lightweight and lightweight concrete due to the low density of rubber aggregate compared to conventional aggregate. In addition, involving waste rubber in construction promotes the development of eco-friendly buildings and encourages the concept of sustainable production.5 However, increasing the rubber content has a negative effect on the compressive strength, tensile strength, flexural strength, and modulus of elasticity.6,7 This can be

ACI Materials Journal/March-April 2016

related to the weak bonding between the rubber particles and surrounding mortar.8 Najim and Hall1 presented a simple investigation for intermediate-scale reinforced concrete beams containing crumb rubber (CR). Eight reinforced concrete beams, two for each mixture—vibrated concrete, vibrated rubberized concrete (VRC), self-consolidating concrete (SCC), and self-consolidating rubberized concrete (SCRC)—were cast with dimensions of 1700 x 200 x 100 mm (66.93 x 7.87 x 3.94 in). The CR replacement reached up to 14% and 18% of the total aggregate volume for VRC and SCRC, respectively. The authors reported that adding CR decreased the flexural capacity and stiffness of beams. Meanwhile, the deformation capacity and energy absorption were increased with increased percentages of CR. Ganesan et al.5 also studied the behavior of SCRC beam-column joints under monotonic and cyclic load. Shredded rubber aggregates were used to replace 15% of the fine aggregate by volume. Their results indicated that the addition of shredded rubber improves the beam-column joint behavior in terms of the energy absorption capacity, crack resistance, and ductility. Meanwhile, SCRC specimens showed a slight reduction in load-carrying capacity. The same behavior was observed in the study conducted by Sadek and El-Attar,9 in which the structural behavior of masonry walls made from rubber-cement bricks was tested. In the production of the bricks, two sizes of rubber were used to replace the coarse and fine aggregates with replacements ranging from 0 to 100% and from 0 to 50% (by volume), respectively. The development of SCRC offers many advantages such as increasing the productivity rate and decreasing the required labor (as it can spread and fill the formwork under its own weight without applying vibration). SCRC also has enough flowability and filling ability to fix the problems of concrete flowing through congested reinforcements. The mixture proportions and components can have some effects on the properties of SCRC. The amount of fine materials (binder) and the percentage of air entrainment in the mixture can affect the mechanical properties of SCRC mixtures. The use of air entrainment can improve the fresh properties10 of the mixture, but will negatively impact mechanical properties. On the other hand, increasing the binder content has ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-166, doi: 10.14359/51688640, received May 22, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

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shown to improve the fresh and mechanical properties of the mixture.11 Using supplementary cementitious materials (SCMs) is one of the ways of potentially enhancing the fresh and mechanical properties of SCRC. Metakaolin (MK) is one of the most effective SCMs that can be used in SCRC and is proven to enhance the mechanical and durability performance of SCRC. Madandoust and Mousavi12 reported that the compressive strength and tensile strength of SCC containing MK were significantly improved (by 27% and 11.1%, respectively) compared to the control mixtures of SCC. Hassan and Mayo13 also observed that the inclusion of 20% MK increased the 28-day compressive strength by 30%. The review of literature indicates that small-scale specimens such as cubes, cylinders, and prisms were used extensively to investigate the performance of rubberized concrete. On the other hand, full-scale testing to study the applicability of this type of concrete for structural applications is significantly lacking, especially when SCRC is used. The main objective of this research was to study the structural performance of full-scale reinforced SCRC and VRC beams under flexural load. A number of SCRC mixtures containing maximum percentages of CR (by volume of fine aggregate) and acceptable fresh properties were developed to cast SCRC beams. Also, additional beams made with VRC mixtures containing maximized percentages of CR were tested for comparison. The investigation included evaluations of the effect of CR on the flexural capacity, cracking behavior, load-deflection response, concrete strain/stiffness, ductility, and toughness of the tested beams. The beams’ mixtures were developed with variable percentages of CR (0 to 50%) using different binder content, the addition of MK, and/or using air entrainment. The investigation also discussed the performance of some code-based equations in predicting the ultimate flexural capacity of the tested beams. RESEARCH SIGNIFICANCE Waste rubber is used in concrete to enhance the ductility, toughness, and impact resistance and reduce the unit weight of the structural members. In addition, using waste rubber in construction promotes the development of eco-friendly concrete and encourages the concept of sustainable production, which is receiving greater attention nowadays. Although there is a growing need to use waste rubber in structural concrete applications, there is a lack of data available regarding the performance of full-scale rubberized concrete elements, especially when SCRC is used. Therefore, this study was conducted to investigate the structural performance of SCRC in full-scale beams. The paper provides information regarding stiffness, ductility, toughness, and cracking behavior of SCRC and VRC beams under flexural load. The authors believe that this investigation will strongly contribute to evaluating the effectiveness of SCRC in structural applications. EXPERIMENTAL PROGRAM Materials properties MK was delivered from the eastern United States, conforming to ASTM C618 Class N.14 The cement used 208

Table 1—Chemical and physical properties of all SCMs used Chemical properties, %

Cement

MK

SiO2

19.64

51 to 53

Al2O3

5.48

42 to 44

Fe2O3

2.38

<2.2

FeO

—

—

TiO2

—

<3.0

C

—

—

Cr2O3

—

—

MnO

—

—

P2O5

—

<0.2

SrO

—

—

BaO

—

—

SO4

—

<0.5

CaO

62.44

<0.2

MgO

2.48

<0.1

Na2O

—

<0.05

C3S

52.34

—

C2S

16.83

—

C3A

10.50

—

C4AF

7.24

—

K2O

—

<0.40

L.O.I

2.05

<0.50

Specific gravity

3.15

2.5

Blaine fineness (m /kg)

410

19,000

2

Note: 1 m2/kg = 4.8824 ft2/lb.

(Type GU) was similar to that of ASTM C618 Type F.14 The chemical and physical properties of cement and MK are shown in Table 1. Natural crushed stone, with a 10 mm (0.394 in.) maximum size, and natural sand were used for the coarse and fine aggregates, respectively. Each aggregate type had a specific gravity of 2.6 and absorption of 1%. A crumb rubber aggregate (with no steel wires) had a maximum size of 4.75 mm (0.187 in.), a specific gravity of 0.95, and negligible absorption was used as a partial replacement of the fine aggregate in SCRC and VRC mixtures. The aggregate gradations of the 10 mm (0.394 in.) crushed stone, natural sand, and CR are presented in Fig. 1. A polycarboxylate-based highrange water-reducer admixture (HRWRA) similar to ASTM C494/C494M15 Type F was used to achieve the required slump flow of SCRC mixtures. An air-entrainment admixture similar to ASTM C260/C260M16 was used to improve the workability of SCRC mixtures. Concrete mixtures A total of 12 concrete mixtures were developed to cast 12 reinforced concrete beams. In general, the experimental investigation aimed to develop a number of SCRC mixtures having maximum percentages of CR (by volume of fine aggregate) and a minimum reduction in strength and stability. To achieve acceptable mixture flowability ACI Materials Journal/March-April 2016

with no sign of segregation in all tested mixtures, a preliminary trial mixtures stage was performed to determine the minimum water-binder ratio (w/b) and the minimum total binder content that can achieve acceptable SCRC flowability without overdosing the HRWRA. The results of the trial mixture stage indicated that at least 0.4 w/b and 500 kg/m3 (31.215 lb/ft3) as a total binder content should be used to obtain SCRC having acceptable slump flow with no visual sign of segregation. Therefore, 0.4 w/b and a minimum of 500 kg/m3 (31.215 lb/ft3) total binder content were used in all tested mixtures (Table 2). Also, a constant coarse-to-fine aggregate ratio (C/F) of 0.7 was chosen for all tested mixtures in this investigation. This ratio was chosen based on previous research13 carried out on SCC with different C/F. During the trial mixtures stage, it was found that the mixtures with 500 kg/m3 (31.215 lb/ft3) binder content and no SCMs (Mixtures 1 to 4) can have a maximum of 15% CR to maintain acceptable SCC fresh properties. Increasing this percentage to 20% resulted in a significant reduction in the passing ability (H2/H1 of L-Box) for all mixtures with 500 kg/m3 (31.215 lb/ft3) binder content. However, when

increasing the total binder content from 500 to 550 kg/m3 (31.215 to 34.335 lb/ft3), the maximum percentage of CR that maintains acceptable SCC fresh properties increased to 20%. The results of the trial mixtures also indicated that using MK enhanced the viscosity of tested mixtures and had a direct impact on improving the particle suspension and passing ability, which allowed a higher percentage (up to 30%) of CR to be used safely in SCRC mixtures. Further increasing the percentage of CR in SCRC mixtures with MK from 30% to 40% required the use of air-entraining admixture (Mixtures  9 and 10) to improve the flowability and passing ability of mixtures. Considering the type of materials used in this investigation, the authors found it very difficult to develop SCRC mixtures with acceptable SCC fresh properties using more than 40% CR. The trial mixtures of this investigation also included developing VRC (Mixtures 11 and 12) to compare its performance with that of SCRC. Because the passing ability and segregation are not factors in VRC mixtures, it was possible to reach a maximum percentage of CR of 50%. Using more than 50% CR in VRC mixtures resulted in a very low compressive strength. The experimental program was divided in two stages. The first stage included four SCRC mixtures with CR percentages varying from 0 to 15% and a binder content of 500 kg/m3 (31.215 lb/ft3). The second stage involved using higher binder content, adding MK and air entrainment, and testing VRC mixtures. The second stage included: 1) two SCRC mixtures with higher binder content of 550 kg/m3 (34.335 lb/ft3) having 15% and 20% CR; 2) two SCRC mixtures with MK having 20% and 30% CR; 3) two SCRC mixtures with MK and air entrainment (0.2205 kg/m3 [0.0138 lb/ft3]) with 30% and 40% CR; and 4) two VRC mixtures with 40% and 50% CR (refer to Table 2). All tested beams were designated by the total binder content, percentage of CR, SCM used, and either the inclusion of micro air (MA) or VRC. For example, a beam containing 550 kg/m3 (34.335 lb/ft3) binder, 40% CR,

Fig. 1—Grading curves for both fine, coarse, and crumb rubber aggregates. (Note: 1 mm = 0.0394 in.) Table 2—Mixture design for tested mixtures Beam no.

Mixture

Cement, kg/m3

SCM (type)

SCM, kg/m3 CA, kg/m3

FA, kg/m3

CR, kg/m3

HRWRA, kg/m3

Density, kg/m3

Stage 1 1

500C-0CR

500

—

—

686.5

980.8

0.0

2.37

2367.3

2

500C-5CR

500

—

—

686.5

931.7

17.9

2.37

2336.2

3

500C-10CR

500

—

—

686.5

882.7

35.8

2.37

2305.1

4

500C-15CR

500

—

—

686.5

833.7

53.8

2.37

2273.9

—

648.1

787.0

50.7

1.84

2255.9

Stage 2 5

550C-15CR

550

—

6

550C-20CR

550

—

—

648.1

740.7

67.7

1.84

2226.5

7

550C-20CR-MK

440

MK

110

638.4

729.6

66.7

5.26

2204.7

8

550C-30CR-MK

440

MK

110

638.4

638.4

100.0

5.26

2146.8

9

550C-30CR-MK-MA

440

MK

110

638.4

638.4

100.0

5.26

2146.8

10

550C-40CR-MK-MA

440

MK

110

638.4

547.2

133.3

5.53

2088.9

11

550C-40CR-MK-VRC

440

MK

110

638.4

547.2

133.3

3.50

2088.9

12

550C-50CR-MK-VRC

440

MK

110

638.4

456.0

166.6

4.00

2031.0

Note: All mixtures have a 0.4 w/b; CA is coarse aggregates; FA is fine aggregates; CR is crumb rubber; 1 kg/m3 = 0.06243 lb/ft3.

ACI Materials Journal/March-April 2016

209

Table 3—Fresh and mechanical properties for tested mixtures Mixture no.

  Slump flow Mixture

Ds, mm

T50, s

V-funnel L-box H2/H1

T0, s

Air, %

28-day fc′

28-day STS

Stage 1 1

500C-0CR

700

1.20

0.89

6.39

1.5

50.2

3.87

2

500C-5CR

690

1.55

0.83

6.95

2.00

43.0

3.23

3

500C-10CR

687

1.74

0.79

7.57

2.3

41.8

2.94

4

500C-15CR

675

2.00

0.75

8.75

4.3

35.3

2.67

5

550C-15CR

710

1.32

0.76

5.97

3.5

37.6

2.73

6

550C-20CR

700

1.54

0.75

6.65

3.2

32.8

2.49

7

550C-20CR-MK

680

2.57

0.86

8.25

3.4

40.8

2.69

8

550C-30CR-MK

620

2.86

0.75

13.5

4.20

34.8

2.36

Stage 2

9

550C-30CR-MK-MA

705

1.53

0.93

5.89

7.5

30.2

2.27

10

550C-40CR-MK-MA

700

1.74

0.84

9.79

8

26.4

1.84

11

550C-40CR-MK-VRC

95

—

—

—

4.5

28.9

2.22

12

550C-50CR-MK-VRC

80

—

—

—

6.1

22.4

1.74

Note: 1 mm = 0.0394 in.

Fig. 2—Dimensions and reinforcement of tested beams. (Note: 1 mm = 0.0394 in.) MK, and MA would be labeled 550C-40CR-MK-MA, and a beam using 550 kg/m3 (34.335 lb/ft3) binder, 50% CR, MK, and VRC would be labeled 550C-50CR- MK-VRC. Casting of beam specimens Twelve full-scale concrete beams were prepared using the 12 developed mixtures. Immediately after mixing, tests on the fresh properties of the concrete mixtures, as well as casting of beams in preassembled wooden forms, were carried out. All SCRC beams were cast without consolidation; the concrete was poured from one side until it flowed and reached the other side. Visual observation showed that the SCRC properly filled the forms with ease of movement around reinforcing bars. On the other hand, VRC beams were consolidated using electrical vibrators and trowel-finished for smooth top surfaces. Formwork was removed after 24 hours of casting, and the beams were moist-cured for 4 days and then air-cured until the date of testing. Fresh and hardened concrete property tests The fresh properties of all tested mixtures were conducted as per the European Guidelines for Self-Compacting Concrete.17 The fresh properties tests included slump flow, V-funnel, and L-box tests. The percentage of air entrainment in the fresh SCC mixtures was measured by following a 210

procedure given in ASTM C231/C231M.18 The compressive strength and splitting tensile strength (STS) tests were conducted using 100 mm (3.94 in.) diameter x 200 mm (7.87 in.) height concrete cylinders, according to ASTM C39/C39M19 and C496/C496M,20 respectively. The compressive strength and STS tests were implemented after the sample had been exposed to condition of curing similar to that of the tested beams. The results of the fresh and mechanical properties of the tested mixtures are presented in Table 3. Flexure test setup, instrumentation, and loading procedure All beams contained shear and flexural reinforcement and were designed to fail in flexure with a ductile behavior. Figure 2 shows the test setup used for all 12 concrete beams during testing. The load was applied through a hydraulic jack (with capacity of 500 kN [112.4 kip]) at a single point and then distributed into two-point loads acting on the beam surface. A linear variable differential transformer (LVDT) and two strain gauges were used to measure the midspan deflection and reinforcement strain, respectively. The strain gauges were installed at the bottom of the longitudinal reinforcement at midspan (maximum flexural moment location). The beams were loaded gradually, with a constant loading rate through four stages until failure (first crack load, and ACI Materials Journal/March-April 2016

50%, 75%, and 100% of the theoretically calculated failure load). After each stage of loading, the cracks were marked and their widths recorded and plotted on each crack pattern. The overall behavior of the beams, including the development of cracks, crack patterns, crack widths, crack heights, and failure modes, was observed and sketched for all beams (Fig. 3). The results obtained from the flexure testing of the 12 tested beams are presented in Tables 4 and 5. DISCUSSION OF TEST RESULTS Fresh properties of SCRC mixtures Table 3 presents the fresh properties of all tested mixtures. In general, as the percentage of CR increased, the fresh properties of SCRC mixtures decreased. The T50 results (the time it takes a mixture to reach 500 mm [19.7 in.] diameter in the slump flow test) and V-funnel time were used to evaluate the viscosity and flowability of SCRC mixtures. The results of Mixtures 1 to 4, which present mixtures with 500 kg/m3 (31.215 lb/ft3) binder and no SCMs, showed that increasing the percentage of CR appeared to increase the mixture viscosity and reduce its flowability. As shown in Table 3, the T50 and V-funnel increased by 66.7% and 36.9%, respectively, as the percentage of CR increased from 0% to 15%. This effect was also found in mixtures with 550 kg/m3 (34.335 lb/ft3) binder content and no SCMs (Mixtures 5 and 6) and mixtures with MK (Mixture 8 compared to Mixture 7, and Mixture 10 compared to Mixture 9), in which the mixture flowability decreased as the CR increased. On the other hand, by comparing Mixture 4 to Mixture 5, it can be observed that increasing the binder content improved the flowability of SCRC and also reduced the dosage of the HRWRA. Meanwhile, by looking at Mixture 6 versus Mixture 7, using MK greatly improved the passing ability (H2/H1 of L-box) of the mixture and caused a reduction in the flowability and a significant increase in HRWRA demand. It should be noted that despite the reduction of the flowability and the increased HRWRA demand of MK mixtures, MK was used in Mixtures 7 to 12 to improve the H2/H1 of L-box to obtain successful SCC passing ability (that is, reach values above 0.75) as per the European Guidelines for Self-Compacting Concrete.17 The result of using higher binder content and/ or adding MK matched other researchers’ results in concrete mixtures without CR.21-23 The results also indicated that adding air entrainment greatly enhanced the mixture flowability (T50 and V-funnel of Mixture 9 compared to Mixture 8). This result also matched other researchers’ results24 where the entrained air in SCC mixtures had a significant effect on improving the mixture flowability. The results of H2/H1 L-box ratio showed that the addition of CR reduced the passing ability of the mixtures. Mixtures 1 to 4 show that increasing the percentage of CR from 0% to 15% reduced the H2/H1 L-box ratio by 15.7%. Using higher binder content (550 kg/m3 [34.335 lb/ft3] instead of 500 kg/m3 [31.215 lb/ft3]) showed a slight enhancement in the passing ability while adding MK increased the L-box ratio significantly, as expected.25 The increase of the passing ability in MK mixtures could be attributed to the fact that the addition of MK improves the mixture viscosity, which contributed to enhancing the distribution and suspenACI Materials Journal/March-April 2016

Fig. 3—Crack patterns of tested beams at failure (crack width in mm). (Note: 1 mm = 0.0394 in.; 1 kN = 0.225 kip.) sion of aggregate particles, and this had a direct impact on improving the passing ability. The addition of air entrainment also showed a significant improvement in the passing ability (Mixture 9 compared to Mixture 8) and secured a higher H2/H1 value, which facilitated developing mixtures with higher CR contents and acceptable passing ability range (above 0.75). This improvement is related to the fact that the air bubbles in concrete mixtures act as a fine aggregate with low surface friction and considerable elasticity, reducing the particle collision/friction and, thus, improving the passing ability.26 The reduction of the passing ability with the increased percentage of CR could be attributed to the 211

Table 4—Results of flexure test Beam no.

Beam ID

First crack load, kN

Failure crack load, kN

At failure Failure type

Number of cracks

Maximum crack width, mm

Stage 1 B1

500C-0CR

32.8

250.0

Flexure

16

5.0

B2

500C-5CR

25.3

251.1

Flexure

18

4.0

B3

500C-10CR

22.8

249.2

Flexure

17

3.5

B4

500C-15CR

21.4

243.3

Flexure

19

3.0

B5

550C-15CR

22.0

246.6

Flexure

14

3.7

B6

550C-20CR

18.2

243.2

Flexure

17

3.3

B7

550C-20CR-MK

20.8

245.0

Flexure

17

3.0

B8

550C-30CR-MK

17.2

228.0

Flexure

16

2.8

Stage 2

B9

550C-30CR-MK-MA

16.5

219.0

Flexure

14

2.5

B10

550C-40CR-MK-MA

13.9

203.6

Flexure

13

2.0

B11

550C-40CR-MK-VRC

14.8

205.7

Flexure

14

2.1

B12

550C-50CR-MK-VRC

14.0

197.5

Flexure

13

2.0

Note: 1 mm = 0.0394 in; and 1 kN = 0.225 kip.

Table 5—Midspan deflection, ductility, and toughness of tested beams Beam no.

Beam ID

Concrete strain at service load (10–6)

Deflection at yield, δy, mm

Deflection at failure, δu, mm

Ductility ratio (δu/δy) Toughness, kN.m

Stage 1 B1

500C-0CR

337.0

10.3

27.0

2.62

4.7

B2

500C-5CR

532.0

9.3

28.5

3.07

5.3

B3

500C-10CR

714.0

8.8

28.2

3.21

5.1

B4

500C-15CR

783.6

9.1

30.8

3.39

5.4

B5

550C-15CR

747.0

8.8

25.9

2.94

4.7

B6

550C-20CR

866.7

8.9

26.8

3.01

5.0

B7

550C-20CR-MK

699.0

9.0

21.9

2.43

3.7

B8

550C-30CR-MK

722.4

9.2

21.3

2.32

3.2

B9

550C-30CR-MK-MA

831.1

8.1

17.9

2.21

2.5

B10

550C-40CR-MK-MA

933.4

9.2

15.7

1.71

1.9

B11

550C-40CR-MK-VRC

980.1

9.0

16.2

1.80

2.0

B12

550C-50CR-MK-VRC

1051.2

9.3

15.9

1.71

1.8

Stage 2

Note: 1 mm = 0.0394 in.; and 1 kN.m = 0.0088 kip.in.

high friction and blocking between crushed stone aggregate and rubber particles. However, all tested mixtures agreed with the limitations given by the European Guidelines for Self-Compacting Concrete17 and the recommended value by the Interim Guidelines for the Use of Self-Consolidating Concrete,27 in which the H2/H1 L-box ratio did not decrease below 0.75. Compressive and splitting tensile strength The 28-day compressive strength and STS of the tested mixtures are shown in Table 3. As seen from Mixtures 1 to 4, increasing the percentage of CR showed a general reduc212

tion in both compressive strength and STS. Varying the CR from 0% to 15% reduced the 28-day compressive strength and STS by 29.6% and 31%, respectively. In Mixtures 5 and 6 (mixtures with 550 kg/m3 [34.335 lb/ft3] binder content), the reduction in the 28-day compressive strength and STS was 12.8% and 13.8%, respectively, as the percentage of CR increased from 15% to 20%. Similar behavior was also noticed in MK mixtures (Mixtures 7 to 12), in which the compressive strength and the STS reduced as the percentage of CR increased. The reduction of the mechanical properties with increased percentages of CR may be attributed to the poor strength of the interfacial transition zone between ACI Materials Journal/March-April 2016

the rubber particles and surrounding mortar, as reported by many researchers.28,29 In addition, the significant difference between the modulus of elasticity of the rubber aggregate and the surrounding mortar can contribute to decreasing the mechanical properties as the CR increased. Moreover, increasing the percentage of CR increased the air content (Table 3), which may also have had a negative effect on the mechanical properties of the mixtures. Increasing the binder content from 500 to 550 kg/m3 (31.215 to 34.335 lb/ft3) raised the compressive strength and STS by 6.5% and 2.25%, respectively, as shown in Mixture 4 compared to Mixture 5. Also, by comparing Mixture 7 to Mixture 6, it can be seen that the addition of MK showed an enhancement in the mechanical properties; the compressive strength and STS increased by 24.4% and 8%, respectively. Meanwhile, from Table 3, using air entrainment helped to develop SCRC with up to 40% CR; however, the 28-day compressive strength and the STS had a reduction of 13.2% and 3.8%, respectively, with the use of air entrainment, as shown in Mixture 8 compared to Mixture 9. The results also indicated that the 28-day compressive strength and STS showed some improvement when using VRC compared to SCRC (Mixture 11 compared to Mixture 10). This can be attributed to the reduction in the air content, as shown in Table 3. It should be noted that the use of VRC (Mixtures 11 and 12) could benefit from using up to 50% CR, in which a further decrease of the mixtures’ self-weight was obtained. Load-deflection characteristics and failure behavior Figure 4 presents the load-central deflection responses of the tested beams. The load and deflection were recorded at the first flexural crack and at various load levels (50, 75, and 100% of failure load). The first flexural cracking load was detected visually and confirmed by the first step or slope change in the load-central deflection response (Fig. 4) and by the load-longitudinal bar strain curves at midspan. Looking closely at Fig. 4, it can be observed that up to the first crack load, the curves appear to be linear with higher stiffness, and then the curves deviate from linearity, showing a reduction in their slopes that indicates lower stiffness due to formation of microcracks. After additional application of load, the longitudinal steels started to yield. During the lifetime between the first crack load and the load that caused steel yielding, the slope of the load-deflection curves changed many times due to multiple cracking. Further increasing the applied load finally caused the concrete crushed in the compression zone and beams to fail. All plots present a typical ductile mode of failure, normally called tension failure, in which the steel bars in tension side yielded before the failure occurrence (as confirmed from the steel strain gauges). The load-deflection curves show that the flexural stiffness (the slope of the load-central deflection curve) of the tested beams decreased as the CR content increased. However, this decrease was not clear in Stage 1 beams (0 to 15% CR) and was more pronounced in beams with higher percentages of CR (more than 20%). This decrease in flexural stiffness is most likely attributed to decreased modulus of elasticity of the SCRC as the CR content increased.1 From Table 5 and Fig. 4, it can ACI Materials Journal/March-April 2016

Fig. 4—Experimental load-midspan deflection responses: (a) Stage 1; and (b) Stage 2. (Note: 1 mm = 0.0394 in.; 1 kN = 0.225 kip.) be observed that increasing the CR content from 0 to 15%, in the first stage, improved the deformation capacity of the tested beams; the maximum deflection increased from 27 to 30.8 mm (1.06 to 1.21 in.). This effect was also noticed up to 20%, as shown in B6 compared to B5 of the second stage. Meanwhile, at high levels of CR replacement (30 to 50%), the rubber-cement composite became weak, which limits the material’s ability to absorb energy and thus exhibits lower deformation capacity. Such behavior proves that using CR up to 20% can enhance the deformation capacity of conventional concrete. Comparing VRC to SCRC (B11 compared to B10) shows that both beams had comparable stiffness and deformation capacity. Concrete strain As mentioned previously, two strain gauges were attached on the top surface of the concrete beams at the midspan. These strain gauges monitored the concrete strain along the history of the beam’s loading (Fig. 5). During the final stage of loading, and before reaching the ultimate failure load, the top surface of the concrete beams was cracked and crushed near the glued strain gauges. Therefore, it was not possible to obtain reliable results from the concrete strain gauges at the ultimate failure load. For this reason, the concrete strain readings in Fig. 5 were recorded up to approximately 95% of the failure load, and therefore the ultimate strain values were higher than the values presented in the figure. It can be generally observed that the slope of the load-strain curve decreased as the percentage of CR increased (in both Stage 1 213

Fig. 5—Experimental load-strain curve of concrete: (a) Stage 1; and (b) Stage 2. (Note: 1 kN = 0.225 kip.) and Stage 2). This result indicates that the concrete stiffness decreased as the percentage of CR increased from 0 to 50%. To focus on the effect of CR specifically at service condition, the values of the concrete strain were recorded at 40% of the ultimate failure load as the customary level service load.30 Table 5 presents the results of the concrete strain at service load for all tested beams. The results showed that varying the percentage of CR from 0 to 15% (Stage 1) raised the strain at service load from 337 × 10–6 to 783.6 × 10–6, respectively. A similar trend of results was noticed in the tested beams of Stage 2. The concrete strain at service load continued to increase with increased CR content. The maximum value of the strain at service load occurred with 50% CR and was 1051.2 × 10–6. Such findings could be attributed to the reduction in the stiffness of rubbercement composite, which resulted from the ability of the rubber particles to undergo large elastic deformation under loading. The results of Fig. 5 and Table 5 also indicated that slight differences in concrete stiffness and strain at service load were noted between VRC and SCRC (B11 compared to B10). Ductility and toughness Displacement ductility was also investigated in this study. Table 5 and Fig. 6 present the ductility ratio μ of the tested beams, which was expressed in terms of μ = δu/δy, where δu is the experimental deflection value at peak failure load, and δy is the experimental deflection at steel yielding. In general, increasing the ductility ratio of the structural member indicates its ability to experience large deflections before failure and, thus, provide ample warning to the occurrence of 214

failure. The results of Stage 1 showed that increasing the CR content improved the ductility of concrete; as the percentage of CR increased from 0 to 15% (B1 compared to B4), μ increased by 29.4%. Replacing the conventional aggregate with rubber aggregate, which has lower stiffness, can greatly enhance the flexibility and energy absorption of rubber-cement composite and, thus, increase the ductility of beams. Stage 2 showed that the ductility enhancement continued up to 20% (B5 compared to B6) replacement level. Further increasing the CR content, however, reduced the ductility of the beams. This reduction may be related to the weakened concrete at the compression zone at higher CR percentages due to the poor bonding between the CR and the surrounding mortar, which limited the beams’ ability to experience higher loading beyond the yielding point. It should be noted that Beams B2 to B4 (Stage 1) and B6 (Stage 2) showed a ductility ratio of 3.07 to 3.39 and 3.01, respectively, which indicate a potential ductility for structural members subjected to large displacements, such as sudden forces caused by earthquake.31,32 Because the use of CR contributed to enhancing the ductility of the tested beams, it was expected that this improvement can directly affect the beam’s toughness. Toughness is the property that can express the capacity of a material to absorb energy up to failure. To compare the toughness of tested beams, the ultimate deformation energy was determined by measuring the area under the load-deflection curve up to the failure load. Table 5 and Fig. 7 show the calculated toughness for all tested beams. Examining the load-deflection curves of the first set of beams (B1 to B4), it can be seen that the area enclosed by the load-deflection curve increased as the CR increased, which indicates an improved toughness of rubberized concrete. For example, increasing the percentage of CR from 0 to 15% raised the toughness by 14.9%. The reason for this increase could be attributed to low stiffness of the CR particles that impart relatively high flexibility and, hence, absorb considerably more energy than could be absorbed by conventional concrete. As shown from the results of Stage 2, the toughness of the tested beams continued to improve up to 20% replacement level (B5 versus B6) and started to drop with higher CR replacement levels (30% to 50%). Toughness is a combination of strength and ductility; the results showed a reduction of the beam’s ductility with a CR percentage higher than 20%. In the meantime, the flexure strength started to drop with increasing the percentages of CR. Therefore, the significant deterioration in the strength and ductility of rubberized concrete at higher levels of CR reduced the ability of concrete to absorb more energy before failure. However, it is worth noting that the possibility of producing SCRC and VRC with higher CR replacement (30% to 50%) contributed to the development of structural members with reduced self-weight. By comparing VRC and SCRC (B10 versus B11), it can be observed that both beams have comparable ductility ratios and toughness, with a slight increase in VRC beams. General cracking and failure behavior As mentioned, the cracks were outlined with a black felt-tipped marker and the crack width was determined and labeled at each loading stage. Figure 3 shows the crack ACI Materials Journal/March-April 2016

Fig. 6—Effect of CR content on ductility: (a) Stage 1; and (b) Stage 2.

Fig. 7—Effect of CR content on toughness: (a) Stage 1; and (b) Stage 2. (Note: 1 kN.m = 0.0088 kip.in.) patterns of all tested beams at the failure stage. During early stages of loading, fine vertical flexural cracks appeared around the midspan of all beams, as expected. With the increase in load, these flexural cracks extended and other new flexural cracks were formed along the loaded span. With a further increase in load (exceeding 50% of theoretical failure load), the flexural cracks that were formed away from the midspan started to propagate diagonally toward the loading points, and other new diagonal cracks began to form separately in locations farther away from the midspan along the beam (Fig. 3). Figure 3 and Table 4 show the crack pattern and crack widths/numbers of all tested beams, respectively. Regarding Stage 1, the beam without CR (B1) appeared to have a larger crack width at failure compared to rubberized concrete beams (B2 to B4). This may be attributed to the higher energy absorption capacity of rubber particles. On the other hand, the failure pattern of rubberized concrete beams (B2 to B4) was characterized by having slightly more cracks than B1. Such results could be related to increasing the midspan deflection (beam’s curvature) as the CR content increased (Table 5), which resulted in the development of more cracks before failure. Increasing the CR content in the second-stage beams also followed the same behavior in terms of higher number of cracks and reduction of the crack width at failure. Table 4 and Fig. 3 also indicated a similar cracking behavior with insignificant differences in terms of crack widths/ numbers for both VRC and SCRC beams (B10 compared to B11).

ACI Materials Journal/March-April 2016

First crack load and ultimate load The first flexural crack load was visually observed and then compared/verified with values associated with the change in slope of the load-deflection and load-longitudinal steel strain curves obtained from the test. Table 4 presents the loads at first flexural crack and failure loads of all tested beams. The results showed that increasing the CR content generally reduced the first crack load and the ultimate failure load in both Stages 1 and 2. Regarding the results of Stage 1, the first crack load appeared to be more affected by increasing the CR content compared to the ultimate failure load, which showed a slight decrease with higher percentages of CR. Increasing the CR content from 0 to 15% reduced the first crack load by 34.76% while the ultimate failure load showed a reduction of 2.67%. The reduction in first crack load could be attributed to the significant deterioration in the tensile strength of the concrete as the CR content increased, as explained previously (results of STS test). Similar behavior was noticed in Stage 2, in which increasing the CR content exhibited a lower first cracking load and lower ultimate failure load. However, the reduction of the ultimate failure load was relatively higher when the percentage of CR exceeded 20%. For example, by comparing B7 to B8, it can be observed that the ultimate failure load reduced by 6.94%. This higher reduction may be due to the decline in the ductility and toughness properties that were noticed at the higher replacement levels (from 30 to 50%). It should be noted that up to 10% CR (Stage 1), the ultimate failure load did not reduce and it only started to drop with higher percentages of CR. This result indicates that using up to 10% CR can help improving the beam’s 215

ductility and toughness (as proved earlier) without affecting the ultimate flexural capacity of the beam. Although using higher percentage of CR (30% to 50%) reduced the ultimate failure capacity of the beams, it contributed to developing semi-lightweight concrete with density varied from 2031 to 2146.8 kg/m3 (126.791 to 134.02 lb/ft3).33 By comparing VRC to SCRC (B10 to B11), it can be observed that both beams showed a comparable ultimate failure load while the first crack load showed a slight increase in VRC beams. This slight increase in the first crack load may be attributed to the improvement of the tensile strength of the VRC mixtures compared to SCRC mixtures. Experimental and theoretical bending moment capacity A comparison between the experimental ultimate moments (Mexp.) and the theoretical design moments (Mtheo.) is shown in Table 6 and Fig. 8. The theoretical design moment of the beams was predicted using the rectangular stress block analysis, as recommended by CSA-0433 and ACI 318-08.34 Table 6—Predictions of ultimate moment capacity Ultimate moment capacity, kN.m

Margin of safety (Mexp./Mtheo.)

Beam no.

Experimental

CSA-04

ACI 318-08

CSA-04

ACI 318-08

B1

85.0

70.7

71.4

1.20

1.19

B2

85.4

69.6

70.2

1.23

1.22

B3

84.7

69.3

69.9

1.22

1.21

B4

82.7

67.9

68.5

1.22

1.21

B5

83.8

68.5

69.1

1.22

1.21

B6

82.6

67.3

67.9

1.23

1.22

B7

83.3

69.1

69.8

1.20

1.19

B8

77.5

67.8

68.4

1.14

1.13

B9

74.5

66.6

67.1

1.12

1.11

B10

69.2

65.4

65.8

1.06

1.05

B11

70.0

66.2

66.7

1.06

1.05

B12

67.2

63.7

64.1

1.05

1.05

Note: 1 kN.m = 0.0088 kip.in.

The comparison showed that the ultimate moments obtained from the experiments were approximately 5 to 23% higher than the predicted values of both CSA-0433and ACI 318-08.34 Increasing the CR content from 0 to 15% (B1 to B4 in Stage 1) and from 15% to 20% (B5 to B6 in Stage 2) showed a slight increase of the value of Mexp./Mtheo., indicating improvement of the flexural capacity of the beams compared to the predicted values. However, increasing the CR content more than 20% showed a general decrease of the value of Mexp./Mtheo.. This finding may be related to the noticeable reduction of the ductility and toughness of the tested beams that occurred at high levels of CR replacement. However, CSA-0433 and ACI 318-0834 can be used to obtain a conservative estimate of the ultimate moment capacity as well as provide an adequate load factor against failure for CR content up to 50%. CONCLUSIONS The structural performance of full-scale reinforced SCRC and VRC beams under flexural load was investigated. The beam mixtures were developed with variable percentages of CR using different binder content, the addition of MK, and/or using air entrainment. The flexural capacity, cracking behavior, load-deflection response, concrete strain/stiffness, ductility, and toughness were studied for all beams. From the results described in this paper, the following conclusions can be drawn: 1. Using CR had an adverse impact on the fresh and mechanical properties of both SCRC and VRC. In SCRC mixtures, the flowability (T50 and V-funnel time), passing ability (H2/H1 of L-box), unit weight, compressive strength, and STS decreased as CR increased while the air content increased. 2. As the percentage of CR increased from 0 to 50%, the first crack load, concrete’s stiffness and beams’ flexural stiffness decreased. On the other hand, the deformation capacity, ductility, and toughness of the tested beams appeared to improve with increases in the CR replacement from 0 to 20% and started to drop with further increases (20 to 50%). 3. No significant difference was noticed between VRC and SCRC beams in terms of their behavior under flexural load. However, at 40% CR, the development of SCRC needed the use of air entrainment to obtain successful SCRC mixtures

Fig. 8—Effect of CR content on predictions of ultimate moment capacity. 216

ACI Materials Journal/March-April 2016

with acceptable passing ability. This essential use of air entrainment in SCRC mixtures resulted in a slight reduction in the compressive strength and STS of SCRC compared to VRC. 4. Using up to 10% CR can improve the beam’s deformation capacity, ductility, and toughness without affecting the ultimate flexural load. However, 10 to 20% CR replacement may continue to improve the beam’s deformation capacity, ductility, and toughness but with a slight reduction in the ultimate flexural load. 5. In this investigation, it was possible to develop SCRC with a maximum CR percentage of 40%. This percentage could be increased to 50% with VRC. However, the 10% increase of CR gave VRC the advantage over SCRC in terms of reducing self-weight while it had a limited advantage in terms of the overall structural behavior of the tested beams. 6. Increasing the percentage of CR more than 20% appeared to affect the conservative estimation for the beams’ moment capacity based on the current ACI 318-08 and CSA-04 design codes. However, ACI 318-08 and CSA-04 can be used to obtain a conservative estimate of the ultimate moment capacity as well as to provide adequate load factor against failure. AUTHOR BIOS

Mohamed K. Ismail is a Graduate Research and Teaching Assistant in the Faculty of Engineering and Applied Science, Memorial University of Newfoundland, St. John’s, NL, Canada. He received his BSc in civil engineering and MSc in structural engineering from Cairo University, Giza, Egypt, in 2010 and 2013, respectively. His research interests include developing and investigating the structural performance of self-consolidating concrete, semi-lightweight and lightweight concrete, and rubberized concrete. Assem A. A. Hassan is an Assistant Professor in the Faculty of Engineering and Applied Science at Memorial University of Newfoundland. His research interests include development and use of self-consolidating, high-performance, and high-strength concretes; rheology of cement paste and concrete mixtures; durability (porosity, diffusivity, chloride permeability) of concrete structures; corrosion monitoring and accelerated corrosion testing of reinforced concrete structures; service life prediction of concrete structures; and performance of large-scale structural members under loading.

ACKNOWLEDGMENTS

The authors would like to acknowledge the Research & Development Corporation of Newfoundland and Labrador (RDC) for sponsoring this work as part of a larger research project.

C/F CR HRWRA MA MK SCC SCM SCRC STS VRC w/b

NOTATION

= coarse-to-fine aggregate = crumb rubber = high-range water-reducer admixture = micro air = metakaolin = self-consolidating concrete = supplementary cementitious material = self-consolidating rubberized concrete = splitting tensile strength = vibrated rubberized concrete = water-binder ratio

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ACI Materials Journal/March-April 2016

ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M21

Tensile Behavior of Steel-Polypropylene Hybrid FiberReinforced Concrete by Lihua Xu, Le Huang, Yin Chi, and Guodong Mei The present study deals with the uniaxial tensile behavior of steel-polypropylene hybrid fiber-reinforced concrete (HFRC). The tensile strengths and complete stress-strain responses of HFRC were measured in terms of different volume fraction and aspect ratio. It was observed that the uniaxial tensile behavior of plain concrete can be significantly improved upon with the addition of hybrid fibers. The steel fiber primarily increases the peak tensile strength, while the polypropylene fiber mainly contributes to increasing residual strength in post-peak response. Subsequently, predictive equations for both the tensile strength and complete stress-strain relation of HFRC were developed, and the results were found in satisfactory agreement with experimental results. Furthermore, a simple elliptic-cap model in tension region was also proposed for the tensile meridian of HFRC within the framework of elastoplasticity, representing a three-dimensional scenario of strength criterion capable of predicting the multiaxial stress state of HFRC in tension region. Keywords: fiber-reinforced concrete; hybrid effect; polypropylene fiber; steel fiber; strength criterion, stress-strain relations; uniaxial tension.

INTRODUCTION In its natural state, concrete is widely applied in engineering applications, owing to essential advantages such as low production cost, formability, and favorable strength and deformation in compression. However, traditional concrete has its disadvantages, such as inherent brittleness, susceptibility to cracking, low tensile strength and deformation threshold, and limited ductility, and at times these disadvantages have been the key determinants, resulting in the selection of other, alternative materials. In an attempt to address some of the aforementioned deficiencies and improve the performance of plain concrete (PC) while reducing the dependency on steel reinforcement, extensive investigations have been undertaken by various researchers over decades. To date, the majority of innovations published have succeeded in using fiber-reinforced concrete (FRC) materials to effectively enhance the composite performance.1,2 With the rapid development of FRC theory and its applications, it becomes well acknowledged that the contribution of fiber is far more apparent when FRC fails in tension rather than in compression.3-6 Because the fibers can provide bridging forces to suppress the crack opening and restrict the crack propagation in the concrete matrix, the tensile properties of the composite can be increased substantially due to the incorporation of fibers. To evaluate the benefits of fiber reinforcement in mechanical performance, considerable experimental efforts have been made over the past few years to investigate the influence of varying fiber types ACI Materials Journal/March-April 2016

and parameters for various loading scenarios.7-14 The corresponding stress-strain responses of both PC and FRC have also been researched in the literature.15-20 However, previous investigations appear to have concentrated on PC, or single FRC with steel fiber-reinforced concrete (SFRC) in particular.16,17,20-24 However, it was found that the mechanical performance of hybrid fiber-reinforced concrete (HFRC) was better balanced, owing to the potential synergistic effects of different fiber combination; both the strength and deformation capacity were significantly improved.3-5,18,25,26 For HFRC, as a relative new cementitious composite, although its mechanical behavior under compression is critical and the corresponding testing method has also been standardized.27,28 However, the tensile properties of HFRC have not been well documented thus far, which can lead to the hesitancy in design of the structural elements within the civil infrastructure when hybrid fibers are involved. The objective of this research is to investigate the mechanical properties of HFRC materials under uniaxial tension. The complete stress-strain responses of HFRC under uniaxial tension were captured and the influences of fiber parameters were analyzed. A preliminary analysis of hybrid effect of different fiber volume fraction on tensile strength was presented. Moreover, predictive equations for the strength and complete stress-strain relation of HFRC under uniaxial tension were developed and an elliptic-cap model for tensile meridian of HFRC was also proposed, which can be properly combined with the model in compression regime for its applicability in finite element simulation of HFRC. RESEARCH SIGNIFICANCE In comparison to single FRC, the mechanical performance can be better balanced through hybrid fiber technology, owing to the potential synergistic effects of different fiber combination. The hybrid fibers act as reinforcement capable of enhancing the matrix performance at a different stage and scale. This paper reports the behavior of steel-polypropylene HFRC under uniaxial tension. The HFRC has shown a favorable combined performance and demonstrated a positive hybrid effect in regard to the tensile strength. This research provides basic tensile properties in precise control and assessment of HFRC composite’s behavior. ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-169.R2, doi: 10.14359/51688641, received June 19, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

219

Table 1—Physical performance of PF and SF Fibers

Type

Length, mm (in.) Diameter, μm (in.) Density, g/cm3 (lb/ft3)

Tensile strength, MPa (ksi)

Elongation rate, %

Aspect ratio

Monofilament PF

SF

Shear corrugated

19 (0.75)

48 (0.0019)

0.91 (56.78)

> 400 (58)

15 to 35

396

13.5, 27, 36 (0.53, 1.06, 1.42)

450 (0.018)

7.8 (486.72)

≥ 600 (87)

—

30, 60, 80

EXPERIMENTAL PROGRAM Materials and mixture proportions The concrete matrix was designed at a 28-day compressive strength of 45 MPa (6.5 ksi). Ordinary portland cement (P.O 42.5) was used as the binder for the mixtures. Local crushed stone with a size between 5 and 20 mm (0.20 and 0.79 in.) was used as coarse aggregate. Normal river sand including 5% of water (by weight) was used as fine aggregate. A highly efficient naphthalene-based water-reducing agent (WRA) with a reducing rate of 20% (reducing water content by 20%) was adopted in the mixture design. The mixture proportion of cement, water, sand, crushed stone, and WRA per cubic meter of the concrete matrix is 441, 150, 794, 1097, and 2.3 kg (742.3, 252.5, 1336.5, 1846.5, and 3.9 lb/yd3), respectively. The mixture design of HFRC was undertaken in accordance to the standard CECS 2004,27 which suggests that, to make full use of the advantage from the improvement of strength and toughness, the volume fracture of steel fiber (SF) should be used between 0.5 and 2.0%, with the aspect ratio (length/diameter) between 30 and 80; and for polypropylene fiber (PF), a low volume fraction, from 0.05 to 0.2%, is recommended, considering the homogeneity and a smooth distribution of PF. Hence, in this study, corrugated SF were used in volume fractions (with respect to the volume of concrete) of 1.1, 1.5, and 1.9% with three aspect ratios of 30, 60, and 80 employed. For PF, early experimental evidence claimed that the aspect ratio of PF has no discernable effect on the variation of mechanical performance of concrete under tension.12,29 Therefore, the PF (trade name of CTA fiber) with a fixed aspect ratio of 396 was chosen throughout the study with volume fractions used in 0.11%, 0.15% and 0.19%, respectively. The corresponding physical properties of fiber texture of both SF and PF are listed in Table 1. To achieve proper dispersion of the hybrid fibers, the coarse aggregates combined with SF were stirred in a mixer for 1 to 2 minutes. The cement and fine aggregates were then added in the mixture. Within 1 to 2 minutes of vibrating, the PF were gradually scattered. Finally, the water and WRA were mixed to complete the mixture preparation. It is noted herein that the mixture slump was controlled at around 80 mm (3.15 in.) by adjusting the dosage of WRA to ensure the workability. Specimen preparation Dog-bone-shaped specimens (a modified shape from the recommendation of CECS 13: 200930 with identical effective cross section and length) were cast in steel molds that 220

Fig. 1—Specimen dimension. were custom-fabricated for the uniaxial tension test. The dimension of the specimen is illustrated in Fig. 1. It has a variable cross section with a total length of 460 mm (18.1 in.). The cross section of both top and bottom part of the specimen is 150 x 150 mm (5.91 x 5.91 in.) with a length of 80 mm (3.15 in.). The cross section of the middle measuring part (colored in gray) is 100 x 100 mm (3.94 x 3.94 in.) with a length of 120 mm (4.72 in.), and the variable cross-sectional segment has a length of 90 mm (3.54 in.), which is curvedesigned with a curvature radius of 174.5 mm (6.87 in.). For such an instance, the cross-sectional area of the measuring part is smaller than that of variable cross-sectional segment, precipitating the occurrence of fracture in the expected gauge length. In addition, a threaded screw with steel wire claws was embedded at each end of the specimen to avoid stress-concentration-induced potential pullout failure. The diameter of the threaded screw is 20 mm (0.79 in.) and the length is 150 mm (5.91 in.) with an embedded depth in the specimen of 80 mm (3.15 in.) (refer to Fig. 2). All the specimens were demolded after 24 hours and stored in a curing room at a temperature of 20 ± 2°C (68 ± 3.6°F) and a relative humidity of 95% until 28-day strength was achieved. ACI Materials Journal/March-April 2016

ment control was applied throughout the testing process with a constant loading rate of 0.04 mm/min (0.0016 in./min). The sampling frequency of the data acquisition system for the measuring device was set to 2 Hz to ensure the synchronization of data collection. Preloading (<3 kN [0.674 kip]) was performed to ensure every joint was smooth and tight.

Fig. 2—Embedded screw with claws. After that, they were ground to a smooth surface and ready for the uniaxial tension test. A total of 102 specimens (34 batches) were tested with respective variables summarized in Table 2, where the nomenclature of specimens is defined as S-A/B/CVs-P-Vp, where S and P stand for the type of fibers (S is SF, and P is PF), A/B/C is the aspect ratio of SF (A = 30, B = 60, and C = 80), and Vs and Vp represent the volume fractions of SF and PF, respectively. For example, the name of SA11P15 represents an HFRC specimen whose volume fractions of SF and PF are 1.1% and 0.15%, respectively, and the aspect ratio of SF is 30. Testing setup The specimens were tested under a specifically designed experimental setup, as shown in Fig. 3. Loading fixtures were gripped on the actuators and connected to the specimen. It contains a spherical joint, protecting the specimen from possible flexural load. It also represents an ideal hinge boundary condition and ensures that the uniaxial stresses are generated in the specimen. To improve the entire stiffness of the testing setup under tensile loading, a rigid frame composed of four dynamometer rods (d = 20 mm [0.79 in.]) and two steel plates was installed around the specimen. The measured stiffness of the four steel dynamometer rods is 122 N/mm (0.7 kip/in.). Ball nuts were used at all the joints of the frame to reduce the potential friction effect. The deformation and the reaction force of the rods were inspected and recorded throughout testing, which were taken into account when the strength and deformation capacity of specimen are evaluated. Two linear variable differential transformers (LVDTs) were mounted at the two opposite lateral sides of specimen to capture the deformation of the measuring section. Strain gauges of size 100 x 3 mm (3.94 x 0.12 in.), also attached at the lateral, were used to monitor the surface cracking. The uniaxial tensile test was then performed using a universal testing machine. Displace-

ACI Materials Journal/March-April 2016

RESULTS AND DISCUSSIONS Typical stress-strain responses Representative stress-strain curves of HFRC in comparison to single SFRC and single polypropylene fiber-reinforced concrete (PFRC) are plotted in Fig. 4. Every stress-strain curve represents the averaged behavior from three specimens. The averaged curve representing the integral trend of the experimental curves was calculated according to Reference 31. The advantages of hybrid fiber, clearly demonstrated in Fig. 4, are that the elastic limit, peak stress, and deformation capacity can be significantly improved. A rough explanation is that the randomly distributed hybrid fibers form a spatial network in the matrix, bridging the cracks and gaps, carrying the tensile forces at different scales during the whole loading process and dissipating energy, consequently resisting the crack opening and propagation until the fibers are pulled out or rupture. However, unlike HFRC, the PC failed in an extremely brittle manner; a sudden fracture was observed during the uniaxial tension test with no valid post-peak response captured. The crack pattern is discussed in the “Reinforcing mechanism” section further in the paper. Furthermore, in all cases shown, the addition of SF can significantly improve both the peak and residual tensile strengths of HFRC (refer to Fig. 4(a)), as well as the pre-peak stiffness slightly. The advantage of adding PF is clearly demonstrated, as the post-peak behavior obviously improved with increasing volume fraction (Fig. 4(b)). In spite of a certain strength reduction occurring occasionally, the rate of strength loss in the post-peak region gradually reduced with the increasing amount of PF, leading to the further increase in residual strength at failure state. This phenomenon is mainly attributed to the fact that the pullout behavior of SF from the matrix plays a dominant role in bridging the macrocrack throughout the tension, in particular at the post-cracking stage when the width of crack becomes larger. The pullout process of SF can help to hinder the crack propagation and delay the failure. In this way, the strength can be significantly improved. However, the PF mainly contributes to the energy absorption during crack development owing to its appealing elongation characteristics upon which it persists to transfer the stress and dissipate energy even under large deformation, such that an expected improvement in post-peak ductility is then observed. This is evidenced by the index Φpost in Table 2 ( Φ post = ∫ε1 f (ε )d ε , where ε1 is uniformly set as 3000 × 10–6), which quantitatively indicates the post-peak energy dissipation capacity to some extent. Consequently, it can be concluded that the behavior of HFRC is in fact a combined behavior of SFRC and PFRC, benefitting from the advantages of both SF and PF in terms of strength, deformation capacity, and post-peak ductility.

221

Table 2—Uniaxial tensile strengths of HFRC No.

Specimens

Vs, %

Aspect ratio of SF

Vp, %

Average tensile strength fft, MPa (ksi)

Percentage of strength increase, %

Average peak strain εft (10–6)

Index Φpost, MPa × 10–6 (ksi × 10–6)

1

SA11P11

1.1

30

0.11

3.640 (0.53)

23.1

198.0

3413.9 (495.5)

2

SB11P11

1.1

60

0.11

3.996 (0.58)

35.2

212.9

4425.3 (642.3)

3

SC11P11

1.1

80

0.11

4.165 (0.60)

40.9

212.7

5345.3 (775.8)

4

SA11P15

1.1

30

0.15

3.730 (0.54)

26.2

217.7

4413.2 (640.5)

5

SB11P15

1.1

60

0.15

4.097 (0.60)

38.6

194.5

4995.7 (725.1)

6

SC11P15

1.1

80

0.15

4.221 (0.61)

42.8

205.5

6565.9 (953.0)

7

SA11P19

1.1

30

0.19

3.849 (0.56)

30.2

198.8

5035.9 (730.9)

8

SB11P19

1.1

60

0.19

4.177 (0.61)

41.3

206.9

5210.4 (756.2)

9

SC11P19

1.1

80

0.19

4.344 (0.63)

46.9

215.3

7699.6 (1117.5)

10

SA15P11

1.5

30

0.11

3.746 (0.54)

26.7

225.8

4638.6 (673.2)

11

SB15P11

1.5

60

0.11

4.228 (0.61)

43.0

212.8

5412.2 (785.5)

12

SC15P11

1.5

80

0.11

4.579 (0.66)

54.9

246.4

8205.2 (1190.9)

13

SA15P15

1.5

30

0.15

4.049 (0.59)

36.9

222.9

5687.6 (825.5)

14

SB15P15

1.5

60

0.15

4.524 (0.66)

53.0

231.8

6710.5 (973.9)

15

SC15P15

1.5

80

0.15

4.694 (0.68)

58.8

259.6

7950.0 (1153.8)

16

SA15P19

1.5

30

0.19

4.087 (0.59)

38.2

212.1

6500.7 (943.5)

17

SB15P19

1.5

60

0.19

4.509 (0.65)

52.5

222.6

7126.7 (1034.4)

18

SC15P19

1.5

80

0.19

4.884 (0.71)

65.2

242.5

9186.6 (1333.3)

19

SA19P11

1.9

30

0.11

3.909 (0.57)

32.2

225.0

5153.4 (748.0)

20

SB19P11

1.9

60

0.11

4.499 (0.65)

52.2

239.3

6943.0 (1007.7)

21

SC19P11

1.9

80

0.11

5.139 (0.75)

73.8

252.1

7753.3 (1125.3)

22

SA19P15

1.9

30

0.15

4.122 (0.60)

39.4

223.8

6382.7 (926.4)

23

SB19P15

1.9

60

0.15

4.384 (0.64)

48.3

249.8

7794.9 (1131.3)

24

SC19P15

1.9

80

0.15

5.490 (0.80)

85.7

271.3

8591.3 (1246.9)

25

SA19P19

1.9

30

0.19

4.917 (0.71)

66.3

235.9

7663.6 (1112.3)

26

SB19P19

1.9

60

0.19

4.511 (0.65)

52.6

269.5

8186.7 (1188.2)

27

SC19P19

1.9

80

0.19

5.238 (0.76)

77.2

279.6

9013.5 (1308.2)

28

SB11

1.1

60

—

3.780 (0.55)

27.9

181.9

3338.7 (484.6)

29

SB15

1.5

60

—

4.158 (0.60)

40.6

203.0

4412.5 (640.4)

30

SB19

1.9

60

—

4.378 (0.63)

48.1

222.3

5495.1 (797.5)

31

P11

—

—

0.11

3.222 (0.47)

9.0

139.1

255.5 (37.1)

32

P15

—

—

0.15

3.185 (0.46)

7.7

142.4

258.7 (37.5)

33

P19

—

—

0.19

3.184 (0.46)

7.7

148.7

297.7 (43.2)

34

PC

—

—

—

2.957 (0.43)

0.0

123.6

192.7 (28.0)

Tensile strength Table 2 summarizes the test results showing the peak stresses and peak strains of HFRC with various fiber parameters as well as those of PC and single FRC subjected to uniaxial tension. All the listed data are the averaged value from three test specimens. It is observed from Table 2 that the peak tensile strengths can be invariably improved from the inclusion of hybrid fibers. The percentage of increase ranges from approximately 25% up to 80% with varying fiber combinations. It is also noted that the main contributor to the substantial increase in uniaxial tensile strength is SF, while PF is found 222

to have insignificant improvement on the tensile strength. In the following sections, particular emphasis is placed on the analysis of the influence of fiber parameters on the tensile strength. Influence of SF Figures 5(a), (b), and (c) show the relationship between the uniaxial tensile strength and the volume fraction of SF (Vs) at polypropylene fiber volume fractions of 0.11%, 0.15%, and 0.19%, respectively, where the aspect ratio of SF in each comparison stays constant throughout. It can be observed from the figures that the tensile strengths gradually increase with increasing steel fiber volume fracACI Materials Journal/March-April 2016

Fig. 3—Experimental setup for uniaxial tension: on-site testing (photograph courtesy of Mei G.D).

Fig. 4—Representative stress-strain curves of HFRC.

Fig. 5—Relationship between uniaxial tensile strength and volume fraction of SF.

tion, and an approximate linear relationship is observed. The physical reason for this phenomenon is that the significant increase in the peak strength of SFRC observed in the macro-scale experiment is due to the contribution of meso-

scale pullout response of SF. The pullout response with different fiber orientation and content can help delay the crack propagation and dissipate energy, which eventually governs the composites tensile behavior.32,33 Even though an

ACI Materials Journal/March-April 2016

223

Fig. 6—Relationship between uniaxial tensile strength and aspect ratio of SF. occasional decrease in strength of Specimen SB19P15 (refer to Fig. 5(b)) is noticed, where high aspect ratio is combined with high volume fraction of both SF and PF, the strength degradation is mainly attributed to the difficulties in proper dispersion of an excessive amount of fibers. It is reasonable that a combination of high volume fraction and high aspect ratio is likely to generate more inherent defects and weaknesses as a result of the voids and microcracks. This is thought to eventually lead to the strength degradation. In addition, Fig. 6 demonstrates the influence of steel fiber aspect ratio on uniaxial tensile strength. The aspect ratio of SF has also demonstrated a significant effect on the improvement of tensile strengths of HFRC samples. For almost all of the cases, a higher aspect ratio results in a higher tensile strength. As mentioned previously and once again proved by the results, the uniaxial tensile strength of HFRC increases with the increase in both steel fiber volume fraction and aspect ratio, and a maximum increase in the tensile strength of 85.7% (Specimen SC19P15) in comparison to PC is observed from the experimental data points. This observation agrees with the previous findings18,26,29 that SF is the primary determinant on improving the strength.

it is noted that tensile strength is also associated with a hybrid effect stemming from various combinations of fiber parameters. The hybrid effect thereof may practically result in a varying trend of improvement of tensile strength. Either a “positive” or a “negative” hybrid effect can be found when the overall performance is compared to the performance evaluated by a simple superposition of individual fiber effect.25 In other words, as previously stated, because effect of PF can be ignored during the analytical formulation of tensile strength of HFRC, the strength of HFRC seems to solely depend on the SF; however, there is a strong indication in Table 2 that, in comparison to the tensile strength of single SFRC, the strength of HFRC with same steel fiber parameter is higher in the majority of cases shown. Hereby, a preliminary qualitative analysis on the influence of various combinations of fiber parameters is performed by defining a hybrid effect coefficient γH35

Influence of PF Figures 7(a), (b), and(c) illustrate the relationship between uniaxial tensile strength and the volume fraction of PF (Vp) at the specified volume fractions of 1.1, 1.5, 1.9% of SF, respectively, where the aspect ratio of SF in each figure is kept constant throughout. No apparent improvement in overall tensile strength associated with the increase of polypropylene fiber volume fraction is observable from the plots. The plots show strength peaks and troughs with no clear indication of the effect of increasing the fiber’s volume fraction. In some cases (refer to Fig. 7(c)), the strengths are noted to actually decrease with increasing volume fraction. Some other published works12,34 also reported that the addition of PF has no discernable impact on improving the uniaxial tensile strength. Therefore, the effect of PF is not taken into account individually as SF when developing the mathematical formulation of tensile strength of HFRC in the following section.

where fs, fp, fft, and fmt denote the uniaxial tensile strength of SFRC, PFRC, HFRC and PC, respectively; and Vtot = Vs + Vp is the total fiber volume fraction. Thereafter, one can easily understand from Eq. (1) that if γH > 1, the hybrid effect is considered to be a positive hybrid effect, otherwise if γH < 1, it is then regarded as a negative hybrid effect. The values of hybrid effect coefficient γH with different fiber volume fraction combinations are shown in Table 3, where the aspect ratio of SF is fixed at 60 (SB series). It is clearly illustrated that all the values of γH are greater than 1, indicating an excellent synergistic effect between SF and PF as well as the positive hybrid effects can be mostly expected. A rough interpretation is shown as follows: because the pullout process of SF is usually accompanied with extensive microcracking within the surrounding matrix, the PF can be active as bridging mechanism that results in the improvement in the debonding and pullout behavior as well as the overall tensile properties. In addition, the hybrid effect coefficient for tensile strength seems to generally increase with the increasing volume fraction of PF, while it basically decreases with an increase of volume fraction of SF. The hybrid effect coefficient of Specimen SB11P19 gives the

Preliminary analysis of hybrid effect Of the limited research available on the tensile performance of HFRC along with the present experimental study, 224



γH =

f ft − f mt

Vp V ( f s − f mt ) s + ( f p − f mt ) Vtot Vtot

(1)

ACI Materials Journal/March-April 2016

Table 3—Hybrid effect of fiber volume fraction on tensile strength γH

Vs, %

Vp, %

Fig. 7—Relationship between uniaxial tensile strength and volume fraction of PF. highest value of 1.659 with a 1.1% volume fraction of SF and a 0.19% volume fraction of PF. Reinforcing mechanism The crack patterns and failure modes were observed for the test specimens. Generally, the failure modes observed in the FRC specimens under uniaxial tension are saliently different ACI Materials Journal/March-April 2016

0

1.1

1.5

1.9

0

—

1

1

1

0.11

1

1.345

1.118

1.136

0.15

1

1.517

1.408

1.070

0.19

1

1.659

1.422

1.184

from those in PC. Figure 8 representatively shows the crack morphology of both PC and FRCs. To illustrate the crack patterns and the fiber’s role in inhabiting crack propagation more clearly, the pictures of the local region (approximately half of the full cross section) were exhibited. From them, it is evident that, for PC, a single major macrocrack traverses through the specimen, a sudden fracture occurred, and the fracture plane is found in a direction nearly perpendicular to the axial loading direction. For PFRC, it is very clear that the failure pattern is quite similar to that of PC. However, the fracture surfaces were observed to be connected by the PF with some of the fibers ruptured. For SFRC, during the loading process, one major crack was formed owing to the propagation and coalescence of microcracks, which is accompanied by the debonding and pullout of SF. The residual stress was mainly carried by the SF without occurrence of specimen fracture. For HFRC, the failure of the specimens seems to be induced by formation of multiple cracks in both macro- and micro-levels with different inclinations; no fiber exposure and concrete spalling were observed. The previously illustrated phenomenon is mainly attributed to the following: for FRC, once the nucleation of microcrack in the matrix is either triggered by microscopic defects or initiated at the weakest link due to the increasing tensile loading, the propagation of microcrack can be effectively restricted by the randomly dispersed hybrid fibers in the matrix. As the load increases, the microcrack develops and runs into a spatial fiber network, where the fibers help to bridge the cracks, carry the tensile stresses, and provide resistance to the further crack opening. Thereafter, the crack encountering the fibers is then branched out or shifted to the other weakest links, resulting in a formation of a set of irregular inclined cracks. As opposed to the failure mode of PC, where a brittle fracture is observed with the specimen sliced into two parts by a main crack, the non-uniformly distributed hybrid fibers can effectively hinder the growth of cracks through an arrest mechanism, dissipating the fracture energy, which significantly improves the matrix fracture toughness. The potential role of fiber in the growth of microcracks was also reported36 in that the microcrack propagation can be arrested by the fiber, resulting in crack bifurcation. Those crack branches further develop and coalesce during loading, which eventually leads to the occurrence of a combination of multiple inclined micro- and macro-cracks. Analytical formulations On the basis of the aforementioned tested data, an equation to calculate the uniaxial tensile strength was developed. 225

Fig. 8—Fracture/crack morphology of PC and FRCs. For the sake of simplicity, a linear relationship between the tensile strength and the fiber reinforcement index is assumed with a term considering a positive hybrid effect added in the equation:

fft = fmt(1 + αλs + βλsλp)­

(2)

where the coefficient α was determined as α = 0.379. Because the positive hybrid effect varies with different fiber combinations of SF and PF, for simplicity, the coefficient β = 0.02 is suggested in this study for a conservative estimation of the hybrid effect. For other hybrid fiber combinations in particular applications with various fiber types or scales, these coefficients can be calibrated through their specific test results. In addition, the parameters fft and fmt denote the uniaxial tensile strength of HFRC and PC, respectively. λs is the steel fiber reinforcement index, calculated as λs = Vs × ls/ds. Vs is the volume fraction of SF, and ls/ds is the aspect ratio of SF. λp is the polypropylene fiber reinforcement index, calculated as λp = Vp × lp/dp. Vp is the volume fraction of PF, and lp/dp is the aspect ratio of PF. Furthermore, a phenomenological constitutive model for predicting the uniaxial tensile behavior of HFRC is presented below. It contains a damage parameter dt, of which the evolution involves a parabola function37 and consists of a curvilinear ascending portion and a descending portion

σ = (1 – dt)Eftε

(3)

1 − ρt [α1 + (1.5 − 1.25α1 ) x + (0.25α1 − 0.5) x 5 ] x ≤ 1  (4) dt =  ρt x >1 1 − α ( x − 1)1.7 + x t 



x=

f ε , ρt = mt (5) ε ft E ft ε ft

where εft is the strain at peak stress that is regressed according to the results in Table 2 as: εft = εt(1 + 0.498λs + 0.697λp), and fmt is the uniaxial tensile strength of concrete matrix. The physical meaning of α1 is the ratio of initial elastic modulus (Eft) and tangential modulus at peak stress (Ep) under tension (that is, α1 = Eft/Ep), which is developed empirically as function fiber reinforcement indexes of hybrid fiber by fitting the experimental curves, given as 226



α1 = 1.2(1 + 0.265λs + 0.277λp) (6)

The parameter αt determines the slope of the descending branch. It is also associated with the fiber reinforcement indexes of hybrid fibers

αt =

0.312 f mt 2 (7) 1 + 3.366λ s + 3.858λ p

It is noted that the parameters α1 and αt respectively lie within the range of 1.2 ≤ α1 ≤ 2 and 0 < αt ≤ 1.5 in this study. For PC, Eq. (6) and (7) reduce to α1 = 1.2 and αt = 0.312fmt2(λs = λp = 0), which is consistent with the equation suggested in literature.38 Figure 9 shows the typical comparisons of experimental stress-strain curves and the fitted curves from predictive equations. The model parameters are respectively calibrated as α1 = 1.813 and αt = 0.352 for SB19P19, and α1 = 1.45 and αt = 0.72 for SA11P11. It is evident that, despite certain fluctuations in the prediction of the strength degradations in post-peak region that are mainly sourced from the regressed Eq. (6) and (7), the predictions can still provide fairly good estimation of the stress-strain behavior of HFRC. Elliptic-cap model In an attempt to predict the multiaxial stress state of HFRC in tension region, a strength criterion was generalized according to the theory of elastoplasticity. Based on the one-dimensional uniaxial tensile strengths, a simple elliptic-cap model in the tension region (hydrostatic pressure ξ > 0) was proposed in this study, which represents a three-dimensional scenario. The aim is to describe the tensile meridian of HFRC, accounting for the notable effect on the addition of hybrid fibers. It was developed in conjunction with the plasticity model of HFRC in compression regime (ξ < 0) proposed in the literature.4 The mathematical formulation of the elliptic-cap model is expressed by the following

f(ρ,ξ) = ρ2 – Aξ2 + Bξ – C = 0

(8)

where the function f(ρ,ξ) defines the shape of meridian that binds the ultimate strength of HFRC in the tension region. The equation is expressed in terms of Haigh-Westergaard coordinates ξ and ρ, where ξ = I1/√3 and ρ = 2 J 2 (I1 is the ACI Materials Journal/March-April 2016

Fig. 10—Boundary conditions. the boundary condition equations that ρthf defines the tensile meridian of HFRC in compression region and yields 2



Fig. 9—Typical experimental stress-strain curves are compared with predictions. first invariant of stress tensor, and J2 is the second invariant of deviatoric stress tensor). A, B, and C are material constants, which can be determined in accordance with three boundary conditions, formulated as

ξ=



ρ



∂ρ ∂ξ

f ft 3

, ρ=

(ξ = 0)

= ρthf

(ξ = 0)

=

∂ρthf ∂ξ

2 f ft (9) 3 (ξ = 0) (10)

(ξ = 0) (11)

These three boundary conditions respectively indicate that: 1) the failure envelope passes through the HFRC’s stress value at failure (that is, the uniaxial tensile strength of HFRC); 2) the cap envelope has a connecting point with the envelope in compression region; and 3) the gradient of tangential line for both envelopes at the connecting point should be identical (refer to Fig. 10). It is also noted from

ACI Materials Journal/March-April 2016

 ρhf   ρhf  ξ = a2  t  + a1  t  + a0 (12) f cu  f cu   f cu 

where ρthf = ktρt, and kt = 1 + 0.08λs + 0.132λp is a reinforcement coefficient of hybrid fibers. More explanations of the parameters in tensile meridian (Eq. (12)) are detailed in the literature.4 Consequently, the coefficients A, B, and C of the cap model can be explicitly calibrated according to the three boundary conditions, and are given by 2



A=

 2   f ft   3 f ft  + B   − C 3    f ft   3 

2

(13)



B = 0.16124ktfcu (14)



C = 0.01449(ktfcu)2 (15)

in which fcu denotes the uniaxial compressive strength of plain concrete matrix. Figure 11 shows representative calibrated elliptic-cap envelopes in conjunction with the tensile meridian of HFRC with varying kt values. The Parameters of A, B, and C and the reinforcement coefficient kt were respectively calibrated according to the equations listed above. For PC, kt = 1.0, and parameters A, B, C were calculated as 3.59, 7.26, and 29.34, respectively. For HFRC, the values are respectively calibrated as kt = 1.1, A = 1.2, B = 7.98, C = 35.5 (for SB11P11), and kt = 1.2, A = 0.22, B = 8.71, and C = 42.25 (for SC15P19). Furthermore, the cap envelope is compared to the triaxial tension results reported in the literature,39 where the volume fraction of SF is 1% with an aspect ratio of 50. The test points lying on the tensile meridian of 0-degree Lode angle with 227

4. A simple elliptic-cap model was also proposed within the framework of elastoplasticity, aiming to describe the failure envelope of HFRC in tension regime, which can serve as a reference to predict the stress state of HFRC in other complex loading conditions. AUTHOR BIOS

Lihua Xu is a Professor in the School of Civil Engineering at Wuhan University, Wuhan, China. She received her PhD from Wuhan University of Hydraulic and Electric Engineering, Wuhan, China. Her research interests include structural and seismic engineering, mechanical behavior, and multi-scale modeling of fiber-reinforced concrete (FRC) materials. Le Huang is a PhD Student in the School of Civil Engineering at Wuhan University. His research interests include the experimental investigation of FRC material and the response of concrete-filled steel tube element.

Fig. 11—Calibrated elliptic-cap envelopes. respective stress ratios of 0.25:0.25:1, 0.5:0.5:1, 0.75:0.75:1, and 1:1:1 were added in Fig. 11; it is clearly observed that the model provides a close and more conservative estimation of strength of FRC in multiaxial tensions. It is also worth mentioning that the proposed elliptic-cap model is capable of reflecting the increase of stress in hydrostatic tension that should be usually arise as a result of increasing fiber content. The suggested elliptic-cap model can be properly combined with the model applicable in the compression regime for it to be usable in a finite element simulation of HFRC. However, it has to be noted that the developed elliptic-cap envelope was generalized into a three-dimensional case based on the uniaxial tensile strength, of which the loading path is unique; considering that the experimental data of HFRC in tension region are scarce, the developed model can serve as a reference in prediction of strength of HFRC in other loading path with different Lode angles (for example, tension-tension-compression, tension-tension-tension scenarios). CONCLUSIONS The mechanical behavior of HFRC under uniaxial tension is investigated in this study. On the basis of the study’s analyses, the major conclusions are summarized as follows: 1. Due to the positive synergetic effect on enhancing the concrete matrix at multi-levels, steel-polypropylene hybrid fiber significantly improves the composite’s tensile strength, deformation capability, and post-peak ductility. 2. Tensile strength improvement of HFRC ranges from 25% to 80%, with respect to PC. SF exerts a primary influence in the hybrid system when compared with PF. Meanwhile, the results also indicate that PF can further improve the residual strength in post-peak response. 3. Empirical equations with respect to the tensile strength and constitutive relation were developed to predict the uniaxial tensile behavior of HFRC, which accounted for the notable effects of varying fiber parameter combinations. The approximations were in good agreement with the experimental results.

228

Yin Chi is an Associate Professor in the School of Civil Engineering at Wuhan University. He received his PhD from the University of Nottingham, Nottingham, UK. His research interests include constitutive modeling of FRC materials and numerical simulation of concrete structures. Guodong Mei is a PhD Student in the School of Civil Engineering at Wuhan University. His interests include the constitutive relation of FRC materials and analysis of FRC structural response.

ACKNOWLEDGMENTS

The authors are grateful for the financial support from the National Natural Science Foundation of China (No. 51508425 and No. 51078295); the Natural Science Foundation of Hubei Province, China (No. 2015CFB171); and the Fundamental Research Funds for the Central Universities (No. 2042015kf0003 and No. 2042014kf0010).

REFERENCES

1. Bentur, A., and Mindess, S., Fibre Reinforced Cementitious Composites, second edition, CRC Press, London, UK, 2006, 624 pp. 2. Swamy, R. N., and Barr, B., Fibre Reinforced Cements and Concretes: Recent Developments, CRC Press, New York, 1989, 716 pp. 3. Chi, Y.; Xu, L.; Mei, G.; Hu, N.; and Su, J., “A Unified Failure Envelope for Hybrid Fibre Reinforced Concrete Subjected to True Triaxial Compression,” Composite Structures, V. 109, 2014, pp. 31-40. doi: 10.1016/j.compstruct.2013.10.054 4. Chi, Y.; Xu, L.; and Yu, H., “Plasticity Model for Hybrid Fiber Reinforced Concrete under True Triaxial Compression,” Journal of Engineering Mechanics, ASCE, V. 140, No. 2, 2014, pp. 393-405. doi: 10.1061/(ASCE) EM.1943-7889.0000659 5. Chi, Y.; Xu, L.; and Zhang, Y., “Experimental Study on Hybrid Fiber-Reinforced Concrete Subjected to Uniaxial Compression,” Journal of Materials in Civil Engineering, ASCE, V. 26, No. 2, 2014, pp. 211-218. doi: 10.1061/(ASCE)MT.1943-5533.0000764 6. Seow, P., and Swaddiwudhipong, S., “Failure Surface for Concrete under Multi-axial Loads—A Unified Approach,” Journal of Materials in Civil Engineering, ASCE, V. 17, No. 2, 2005, pp. 219-228. doi: 10.1061/ (ASCE)0899-1561(2005)17:2(219) 7. Atis, C.; Karahan, O.; Ari, K.; Celik Sola, Ö.; and Bilim, C., “Relation between Strength Properties (Flexural and Compressive) and Abrasion Resistance of Fiber (Steel and Polypropylene)-Reinforced Fly Ash Concrete,” Journal of Materials in Civil Engineering, ASCE, V. 21, No. 8, 2009, pp. 402-408. doi: 10.1061/(ASCE)0899-1561(2009)21:8(402) 8. Chern, J.; Yang, H.; and Chen, H., “Behavior of Steel Fiber Reinforced Concrete in Multiaxial Loading,” ACI Materials Journal, V. 89, No. 1, Jan.-Feb. 1992, pp. 32-40. 9. Graybeal, B. A., and Baby, F., “Development of Direct Tension Test Method for Ultra-High-Performance Fiber-Reinforced Concrete,” ACI Materials Journal, V. 110, No. 2, Mar.-Apr. 2013, pp. 177-186. 10. Kawamata, A.; Mihashi, H.; and Fukuyama, H., “Properties of Hybrid Fiber Reinforced Cement-Based Composites,” Journal of Advanced Concrete Technology, V. 1, No. 3, 2003, pp. 283-290. doi: 10.3151/ jact.1.283 11. Lawler, J.; Zampini, D.; and Shah, S., “Microfiber and Macrofiber Hybrid Fiber-Reinforced Concrete,” Journal of Materials in Civil Engineering, ASCE, V. 17, No. 5, 2005, pp. 595-604. doi: 10.1061/ (ASCE)0899-1561(2005)17:5(595) 12. Xu, L.; Xu, H.; Chi, Y.; and Zhang, Y., “Experimental Study on Tensile Strength of Steel Polypropylene Hybrid Fiber Reinforced

ACI Materials Journal/March-April 2016

Concrete,” Advanced Science Letters, V. 4, No. 3, 2011, pp. 911-916. doi: 10.1166/asl.2011.1740 13. Yang, E.; Wang, S.; Yang, Y.; and Li, V. C., “Fiber-Bridging Constitutive Law of Engineered Cementitious Composites,” Journal of Advanced Concrete Technology, V. 6, No. 1, 2008, pp. 181-193. doi: 10.3151/ jact.6.181 14. Yao, W.; Li, J.; and Wu, K., “Mechanical Properties of Hybrid Fiber-Reinforced Concrete at Low Fiber Volume Fraction,” Cement and Concrete Research, V. 33, No. 1, 2003, pp. 27-30. doi: 10.1016/ S0008-8846(02)00913-4 15. Ahmed, S., and Maalej, M., “Tensile Strain Hardening Behaviour of Hybrid Steel Polyethylene Fibre Reinforced Cementitious Composites,” Construction & Building Materials, V. 23, No. 1, 2009, pp. 96-106. doi: 10.1016/j.conbuildmat.2008.01.009 16. Barragán, B.; Gettu, R.; Martin, M.; and Zerbino, R. L., “Uniaxial Tension Test for Steel Fibre Reinforced Concrete—A Parametric Study,” Cement and Concrete Composites, V. 25, No. 7, 2003, pp. 767-777. doi: 10.1016/S0958-9465(02)00096-3 17. Bencardino, F.; Rizzuti, L.; Spadea, G.; and Swamy, R., “StressStrain Behavior of Steel Fiber-Reinforced Concrete in Compression,” Journal of Materials in Civil Engineering, ASCE, V. 20, No. 3, 2008, pp. 255-263. doi: 10.1061/(ASCE)0899-1561(2008)20:3(255) 18. Li, Z.; Li, F.; and Chang, T., “Uniaxial Tensile Behavior of Concrete Reinforced with Randomly Distributed Short Fibers,” ACI Materials Journal, V. 95, No. 5, Sept.-Oct. 1998, pp. 564-572. 19. Park, S.; Kim, D.; Ryu, G. S.; and Koh, K. T., “Tensile Behavior of Ultra High Performance Hybrid Fiber Reinforced Concrete,” Cement and Concrete Composites, V. 34, No. 2, 2012, pp. 172-184. doi: 10.1016/j. cemconcomp.2011.09.009 20. Yang, M.; Huang, C.; and Wang, J., “Characteristics of Stress-strain Curve of High Strength Steel Fiber Reinforced Concrete under Uniaxial Tension,” Journal of Wuhan University of Technology—Materials, Science Editor, V. 21, No. 3, 2006, pp. 132-137. 21. Li, Q., and Ansari, F., “High-Strength Concrete in Uniaxial Tension,” ACI Materials Journal, V. 91, No. 1, Jan.-Feb. 2000, pp. 49-55. 22. Li, V.; Wu, H.; Maalej, M.; Mishra, D. K.; and Hashida, T., “Tensile Behavior of Cement-Based Composites with Random Discontinuous Steel Fibers,” Journal of the American Ceramic Society, V. 79, No. 1, 1996, pp. 74-78. doi: 10.1111/j.1151-2916.1996.tb07882.x 23. Sorelli, L.; Meda, A.; and Plizzari, G., “Bending and Uniaxial Tensile Tests on Concrete Reinforced with Hybrid Steel Fibers,” Journal of Materials in Civil Engineering, ASCE, V. 17, No. 5, 2005, pp. 519-527. doi: 10.1061/(ASCE)0899-1561(2005)17:5(519) 24. Thomas, J., and Ramaswamy, A., “Mechanical Properties of Steel Fiber-Reinforced Concrete,” Journal of Materials in Civil Engineering, ASCE, V. 19, No. 5, 2007, pp. 385-392. doi: 10.1061/ (ASCE)0899-1561(2007)19:5(385)

ACI Materials Journal/March-April 2016

25. Mei, G.; Xu, L.; Li, S.; and Chi, Y., “Hybrid Effects on Strength of Steel-polypropylene Hybrid Fiber Reinforced Concrete under Uniaxial and Triaxial Compression,” Applied Mechanics and Materials, V. 268, 2013, pp. 782-787. 26. Qian, C., and Stroeven, P., “Development of Hybrid Polypropylene-Steel Fibre-Reinforced Concrete,” Cement and Concrete Research, V. 30, No. 1, 2000, pp. 63-69. doi: 10.1016/S0008-8846(99)00202-1 27. CECS, 2004, “Technical Specification for Fiber Reinforced Concrete Structure,” China Planning Press, China, 2004. 28. RILEM TC-162 TDF., “Final Recommendation of RILEM TC 162-TDF: Test and Design Methods for Steel Fibre Reinforced Concrete Sigma-Epsilon-Design Method,” Materials and Structures, V. 36, No. 262, 2003, pp. 560-567. doi: 10.1617/14007 29. Tavakoli, M., “Tensile and Compressive Strengths of Polypropylene Fiber Reinforced Concrete,” Fiber-Reinforced Concrete: Developments and Innovations, SP-142, J. I. Daniel and S. P. Shah, eds., American Concrete Institute, Farmington Hills, MI, 1994, pp. 61-72. 30. CECS 13: 2009, “Standard Test Methods for Fiber Reinforced Concrete,” China Planning Press, China, 2009. 31. Ren, X.; Yang, W.; and Zhou, Y., “Behavior of High-Performance Concrete under Uniaxial and Biaxial Loading,” ACI Materials Journal, V. 105, No. 6, Nov.-Dec. 2008, pp. 548-557. 32. Markovich, I.; Van Mier, J. G. M.; and Walraven, J. C., “Single Fibre Pullout from Hybrid Fiber Reinforced Concrete,” HERON, V. 46, No. 3, 2001, pp. 191-200. 33. Sato, Y.; Van Mier, J. G. M.; and Walraven, J. C., “Mechanical Characteristics of the Multi-Modal Fiber Reinforced Cement Based Composites,” Proceedings of the 5th International RILEM Symposium, BEFIB, 2000, pp. 791-800. 34. Song, P. S.; Hwang, S.; and Sheu, B. C., “Strength Properties of Nylon- and Polypropylene-Fiber-Reinforced Concretes,” Cement and Concrete Research, V. 35, No. 8, 2005, pp. 1546-1550. doi: 10.1016/j. cemconres.2004.06.033 35. Wang, P.; Huang, Z.; and Zhou, D., “Impact Mechanical Properties of Concrete Reinforced with Hybrid Carbon Fibers,” Journal of Vibration and Shock, V. 31, No. 12, 2012, pp. 14-18. 36. Mobasher, B.; Stang, H.; and Shah, S. P., “Microcracking in Fiber Reinforced Concrete,” Cement and Concrete Research, V. 20, No. 5, 1990, pp. 665-676. doi: 10.1016/0008-8846(90)90001-E 37. Guo, Z., Principles of Reinforced Concrete, Butterworth-Heinemann, 2014, 590 pp. 38. China Standardization, “Code for Design of Concrete Structures (GB50010-2010),” China Architecture & Building Press, Beijing, China, 2010. 39. Song, Y.; Zhao, G.; Peng, F.; Huang, C.; Hu, B.; and Shen, J., “Strength Behavior and Failure Criterion of Steel Fibre Concrete under Triaxial Stresses,” China Civil Engineering Journal, V. 27, No. 3, 1994, pp. 14-23.

229

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This collection contains 10 papers selected from the three special sessions held at The Concrete Convention and Exposition in Washington, DC, October 2014. Topics include: Performance Reliability of Reinforced Concrete Beams Strengthened with Fiber-Reinforced Polymer (FRP) Sheets, Reducing Deck Cracking in Composite Bridges by Controlling Long-Term Properties, and many more.

In October 2015, the Italy Chapter – ACI (ACI IC) and the Department of Civil, Chemical, Environmental, and Material Engineering (DICAM) of the University of Bologna sponsored the First International Workshop on “Durability & Sustainability of Concrete Structures” in Bologna, Italy. The proceedings of the workshop consist of 48 papers concerning reduction in greenhouse gases in cement and concrete industry, recycled materials, innovative binders and geopolymers, life-cycle cost assessment in concrete construction, reuse and functional resilience of reinforced concrete structures, repair and maintenance, testing, inspection, and monitoring.

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ACI MATERIALS JOURNAL

TECHNICAL PAPER

Title No. 113-M22

Nano-Modified Fly Ash Concrete: A Repair Option for Concrete Pavements by A. Ghazy, M. T. Bassuoni, and A. Shalaby Efficient repair of concrete pavements typically requires a rapid-setting material to accelerate opening the road to traffic. While numerous high-early-strength cementitious repair materials are commercially available, many of them are vulnerable to premature deterioration. On the other hand, despite its improved long-term performance, concrete incorporating fly ash is rarely used as a repair material due to the delay in setting time, strength gain, and microstructural development at early ages. Nevertheless, these performance limitations can be mitigated by incorporation of nanoparticles (for example, nanosilica) in fly ash concrete. In this study, an effort was made to develop nano-modified fly ash concrete as a repair material for concrete pavements. The performance of the newly developed mixtures was compared to that of two commercial cementitious products. The results indicate that the nano-modified fly ash concrete has balanced performance in terms of hardening time, strength development, bonding with substrate concrete, and resistance to infiltration of fluids and salt-frost scaling. Keywords: durability; early age; fly ash; nanosilica; repair.

INTRODUCTION Premature failure of repairs in concrete pavements and bridge decks is frequently observed, resulting in significant life-cycle, economic, and social losses.1,2 Efficient repair of concrete pavements typically requires a rapid-setting material that can be placed and hardened within a relatively short period of time for quick opening to traffic. While numerous high-early-strength cementitious repair materials are commercially available, many of these materials are vulnerable to cracking, poor bonding, and premature deterioration, for example, due to incompatibility with the existing concrete pavement.2,3 In addition, some studies have shown concerns of using high-early-strength concrete in repair applications for pavements.4 These materials can lead to stress concentrations because of their susceptibility to thermal gradients and autogenous shrinkage, resulting in high levels of microcracking and, in turn, durability issues. Extensive research on the use of supplementary cementitious materials (SCMs) such as fly ash showed that incorporation of Class F fly ash generally improves the long-term performance and durability of concrete.5-7 Despite the benefits of fly ash concrete, practical limitations remain unresolved in field applications. The delay in setting time, strength gain, and microstructural development at early ages of fly ash concrete are considered to be the major issues, which deter its wider acceptance as a repair material.6-8 Also, a number of laboratory studies have indicated inferior scaling resistance of concrete containing dosages of fly ash in excess of 25 to 30% of the binder when subjected to cycles of freezing and thawing in the presence of deicing chemicals.6,7 ACI Materials Journal/March-April 2016

Recently, nanomaterials have been progressively applied in the field of concrete research, attracting considerable scientific interest due to the new potential uses of nanometer-sized particles in cementitious binders. Concrete with superior properties can be produced by incorporating nanoparticles with fly ash.8-11 As a result of their ultrafine nature (size scale of 1-100 billionth of a meter), nanoparticles can vigorously speed up the kinetics of cement hydration in concrete.8-11 Hence, the delay in setting time, strength gain, and microstructural development of fly ash concrete may be mitigated. RESEARCH SIGNIFICANCE Carefully balancing the early-age and long-term performance of cement-based repair materials remains a challenging task, which warrants further investigation. In comparison to two cementitious products specified by transportation agencies in Manitoba for partial-depth repair of concrete pavements, an effort was made in the present study to develop nano-modified fly ash concrete as a repair material for concrete pavements to achieve balanced performance in terms of hardening time, strength development, bonding with substrate concrete, and durability to infiltration of fluids and salt-frost scaling. EXPERIMENTAL PROCEDURE Materials General use (GU) portland cement and fly ash (Class F), which meet the requirements of the CAN/CSA-A300112 standard, were used as the main components of the binder. Their chemical and physical properties are shown in Table 1. In addition, a commercial nano-silica sol (50% solid content of SiO2 dispersed in an aqueous solution [Table 1]) was incorporated in all binders. Six concrete mixtures were prepared and a non-chloride accelerator, complying with ASTM C494/C494M13 Type E, was used in three mixtures to accelerate the setting time. Locally available coarse aggregate (natural gravel with a maximum size of 9.5 mm [0.375 in.]) and fine aggregate (well-graded river sand with a fineness modulus of 2.9) were used. The specific gravity and absorption were 2.65 and 2%, respectively, for gravel, and 2.53 and 1.5%, respectively, for sand. A high-range water-reducing ACI Materials Journal, V. 113, No. 2, March-April 2016. MS No. M-2015-185.R1, doi: 10.14359/51688642, received June 18, 2015, and reviewed under Institute publication policies. Copyright © 2016, American Concrete Institute. All rights reserved, including the making of copies unless permission is obtained from the copyright proprietors. Pertinent discussion including author’s closure, if any, will be published ten months from this journal’s date if the discussion is received within four months of the paper’s print publication.

231

Table 1—Properties of GU cement, fly ash, and nanosilica Cement

Fly ash

Nanosilica

SiO2,%

19.21

55.20

99.17

Al2O3,%

5.01

23.13

0.38

Fe2O3,%

2.33

3.62

0.02

CaO,%

63.22

10.81

—

MgO,%

3.31

1.11

0.21

SO3,%

3.01

0.22

—

Table 2–Composition and properties of commercial products according to manufacturers’ datasheets A Composition, % by mass Hydraulic cement

*

7 to 13

Silica sand, crystalline



>60

Titanium dioxide



0.1 to 1

Borax



1 to 5

Physical properties

Na2Oeq,%

0.12

3.21

0.20

Specific gravity

3.15

2.12

1.40

Mean particle size, µm (in.)

13.15 (8.45 × 10–6)

16.56 (6.52 × 10–6)

35 × 10–3 (1.38 × 10–6)

Water content, L/22.7 kg (gal/50 lb)

Fineness, m2/kg (ft2/lb)

390 (1.90 × 103)*

290 (1.41 × 103)*

80,000 (0.39 × 106)†

Viscosity, Cp (ft·s/lb)

—

—

8 (0.537 × 10–2)

pH

—

—

9.5

Specific gravity

2.7

2.75

2.1 to 2.8 (0.56 to 0.75)

1.6 to 1.8 (0.42 to 0.47)

Mixing time, min

4 to 5

4

Extension

50%

50%

†

Curing procedures

Wet cure surface with water and polyethylene sheets at least 1 day, or use a curing compound

*

Blaine fineness.

*Range is not specified in datasheet.

†

Fineness was determined by titration with sodium hydroxide.

†

admixture (HRWRA) based on polycarboxylic acid and complying with ASTM C494/C494M13 Type F was added to maintain a slump range of 50 to 100 mm (1.97 to 3.94 in.). In addition, an air-entraining admixture was used to provide a fresh air content of 6 ± 1%. For comparison purposes, two commercial cementitious repair materials were evaluated. These products are approved and used in the province of Manitoba based on the premise that they achieve early-age performance and adequate service life. Table 2 shows their composition and physical properties according to each manufacturer’s datasheet. Procedures The formulations of the nano-modified fly ash concrete stem from a compatibility perspective between repair and parent concrete; thus, all the repair mixtures comprised fly ash comparable to concrete pavements in Manitoba, in which 15% fly ash is typically used. Three normal-setting concrete mixtures (designated as N) were prepared with GU cement and variable dosages of fly ash (15%, 22.5%, and 30% replacement by mass of the base binder (385 kg/m3 [24 lb/ft3]) comprising GU cement and fly ash); nanosilica was added to the mixtures at a single dosage of 6% by mass of the base binder (that is, a solid content of 23 kg/m3 [1.44 lb/ft3]). In addition, three corresponding rapid-setting concrete mixtures (designated as R) were produced with an accelerating admixture. For all mixtures, the total cementitious materials (ternary binder: GU cement, fly ash, and nanosilica) content and water cementitious materials ratio (w/cm) were kept constant at 408 kg/m3 (25 lb/ft3) and 0.38, respectively. Constituent materials were mixed in a concrete mixer with a speed of 60 rpm. To attain homogenous dispersion of components, a specific sequence of mixing was adopted based on experimental trials. First, approximately 15% of the mixing water was added to the aggregate while mixing 232

B

Coarse aggregate extension by mass of repair material per bag, 22.7 kg (50 lb).

for 30 seconds. The cement and fly ash were then added to the aggregate and mixed together for 60 seconds. The nanosilica and the admixtures (air-entraining admixture, HRWRA, and accelerator) were added to the remaining water while stirring vigorously for 45 seconds to obtain a liquid phase containing well-dispersed nanoparticles and admixtures. Finally, the liquid phase was added to the mixture and mixing continued for 2 minutes. After mixing and casting the concrete, a vibrating table was used to ensure good compaction of specimens. Polyethylene sheets were used to cover the surface of specimens for 24 hours. The specimens were then demolded and cured in a standard curing room (maintained at a temperature of 22 ± 2°C [72 ± 3.6°F] and a relative humidity of more than 95%) until testing. The proportions of the nano-modified mixtures are shown in Table 3. For the commercial products, the manufacturers’ recommendations were carefully followed in the proportioning, mixing, casting, and curing procedures (Table 2). Testing methods To determine the setting time, the mortar fraction of each mixture (portion passing sieve No. 4 [4.75 mm (0.19 in.)]) was placed in a container at room temperature as specified by ASTM C403.14 At regular time intervals, the penetration resistance was determined by standard needles. In addition, paste samples with identical proportions to the paste fraction in the concrete mixtures (Table 3) were prepared to measure the heat released from the hydration reactions by an isothermal calorimeter kept at 23°C (73°F) following the general guidelines of ASTM C1679.15 The rate of heat generated was monitored and recorded every minute continuously for 100 hours, and it was normalized by the mass of the sample. Also, the cumulative heat released was determined. For each mixture, triplicate 100 x 200 mm (4 x 8 in.) concrete cylinders were prepared for the compressive ACI Materials Journal/March-April 2016

Table 3—Proportions of mixtures per cubic meter of concrete Mixture ID

*

Cement, kg

Fly ash, kg

Nanosilica, kg

Water*, kg

Coarse aggregate, kg

Fine aggregate, kg

HRWRA, L/m3

Accelerator, L/m3

F15

327

58

0

147

858

858

0.7

0

F30

269

116

0

147

850

850

0.4

0

NF15

327

58

46

131

830

830

2.3

0

NF22.5

298

87

46

131

830

830

2.1

0

NF30

269

116

46

131

830

830

1.9

0

RF15

327

58

46

131

830

830

1.8

6.9

RF22.5

298

87

46

131

830

830

1.7

6.9

RF30

269

116

46

131

830

830

1.5

6.9

Adjusted amount of mixing water considering water content of nanosilica (aqueous solution with 50% solid content of SiO2).

Notes: 1 kg = 2.205 lb; 1 L/m3 = 7.48 × 103 gal/ft3.

strength test according to ASTM C3916 and splitting tensile strength test according to ASTM C496,17 which were performed at different ages. To evaluate the bond between the repair mixtures and concrete substrate along with the resistance of the composite assembly to environmental conditioning, the pulloff test was used according to CSA A23.2-6B.18 Concrete slabs of 300 x 400 mm (11.8 x 15.8 in.) surface area and 140 mm (5.5 in.) thickness were used as concrete substrate (350 kg [772 lb] GU cement with 15% fly ash as a binder replacement and 0.38 w/cm). After casting, the slabs were demolded and moist cured for 7 days in the curing room and then maintained in normal laboratory conditions. At 90 days, the top surface (finished surface) was wire brushed and cleaned; subsequently, the repair mixtures were placed on the top surface with a thickness of 80 mm (3.15 in.). After moist curing for 28 days, the slabs were partially cored to determine the pulloff strength according to CSA A23.2-6B.18 Furthermore, the bond strength of companion slabs was evaluated after consecutive freezing-and-thawing (F/T) and wetting-anddrying (W/D) cycles. A total of 25 F/T cycles followed by 25 W/D cycles were applied. This customized procedure simulates climatic conditions of successive winter and summer seasons, which correlates to in-service conditions. For the F/T stage, ASTM C672/C672M19 regime was applied for 25 cycles. Subsequently, the specimens were exposed to 25 W/D cycles, where each cycle consisted of ponding (3 to 5 mm [0.12 to 0.20 in.]) the surface of specimens with 4% calcium chloride solution for 16 hours at a temperature of 22 ± 2°C (73 ± 3.6°F), followed by drying at 40 ± 2°C (104 ± 3.6°F) and 40 ± 5% RH for 8 hours. At 28 days, the resistance of the mixtures to the penetrability of aggressive ions was evaluated by the rapid chloride penetrability test (RCPT) according to ASTM C1202.20 To avoid the electrolysis bias of this method, the penetration depth of chloride ions into concrete, which better correlates to the physical characteristics of the pore structure, was determined according to the procedure described by Bassuoni et al. in 2006.21 After the RCPT, the specimens were axially split and sprayed with 0.1 M silver nitrate solution, which forms a white precipitate of silver chloride in approximately 15 minutes, to measure the physical penetration depth of chloride ions. The average depth of the white precipitate was determined at five different locations along ACI Materials Journal/March-April 2016

the diameter of each half specimen. This depth is considered to be an indication of the ease of ingress of external fluids and, thus, the continuity of the microstructure.21 In addition, ASTM C672/C672M19 was conducted for all the mixtures to evaluate their resistance to surface scaling due to deicing salts (4% calcium chloride) and F/T cycles. The resistance to surface scaling was evaluated when the curing method (by a chemical compound or a standard curing room) and time (3 to 14 days) were varied. Moreover, the resistance to surface scaling was evaluated qualitatively by visual examination, and quantitatively by mass of scaled materials. Thermal and microscopy studies were conducted to evaluate the evolution of microstructure in the concrete mixtures. The quantity of portlandite (calcium hydroxide) in the matrix was determined up to 90 days to assess the effect of nanosilica and fly ash on the hydration and pozzolanic reactions. Thermogravimetry (TG) at a heating rate of 10°C/min (18°F/min) was used for this purpose on powder samples extracted from the concrete mixtures. The content of portlandite was calculated by determining the percentage drop of an ignited mass of the TG curves at a temperature range of 400 to 450°C (752 to 842°F) and multiplying it by 4.11 (ratio of the molecular mass of portlandite to that of water). Backscattered scanning electron microscopy (BSEM) with elemental dispersive X-ray (EDX) were conducted on polished thin sections from the concrete mixtures. At 28 days, slices were cut from specimens, which were then dried and impregnated by a lowviscosity epoxy resin under vacuum pressure and polished by successive diamond surface-grinding to a thickness of 30 to 50 µm (1.18 × 10–3 to 1.97 × 10–3 in.). The sections were coated with carbon to enhance the conductivity for the BSEM analysis. Finally, selected tests (setting time, heat of hydration, and thermal analysis) were also done on reference fly ash concrete mixtures (without nanosilica) comprising 15% and 30% fly ash (F15 and F30) to exemplify the difference in behavior relative to the nano-modified fly ash concrete. EXPERIMENTAL RESULTS AND DISCUSSION Fresh properties and heat of hydration Table 4 shows the properties of fresh concrete, including density, slump, and setting times (initial and final). Products A and B showed very dry consistency (zero slump) and had very short hardening times (40 to 60 minutes), as indi233

Table 4—Density, slump, and initial and final setting times Mixture ID.

A

B

F15

F30

NF15

NF22.5

NF30

RF15

RF22.5

RF30

Density, kg/m

2312

2280

2249

2233

2224

2224

2224

2231

2231

2230

Slump, mm

0

0

75

95

55

65

80

75

90

100

Initial, min

40

40

450

550

270

345

360

165

175

210

Final, min

45

60

605

830

415

475

505

285

295

340

3

Notes: 1 kg/m3 = 0.0624 lb/ft3; 1 mm = 0.0394 in.

Fig. 1—Penetration resistance versus time. (Note: 1 MPa = 145 psi.) cated by the acute increase of the penetration resistance curves (Fig. 1). This may imply critical placement, consolidation, and finishing, as all these procedures should be completed within 40 minutes. Also, when large patches are repaired, there will be a high potential for cold joints as repair mortars/concretes are typically mixed on site in small batches. Comparatively, the nano-modified fly ash concrete mixtures had better workability, especially the R mixtures owing to the effect of Type E (water reducer and accelerator) admixture combined with the HRWRA. Also, they had ample setting times (Fig. 1), which would allow more flexibility in the repair process, especially for large or multiple patches. The effect of higher dosages of fly ash on retarding the setting time of concrete is well-documented,5-7 as depicted in Fig. 1. For example, the reference concrete mixture containing 30% fly ash (F30) exhibited initial and final setting times of, respectively, 100 and 225 minutes longer than that of the reference mixture with 15% fly ash (F15). However, the setting times of the nano-modified mixtures were significantly shortened relative to the reference mixtures. For example, the initial setting times for NF15 and RF15 were 270 and 165 minutes, respectively, compared to 450 minutes for the corresponding reference mixture, F15. Also, at a higher dosage of fly ash, Mixtures NF30 and RF30 exhibited final setting times 40% and 60%, respectively, shorter than the reference mixture F30. This is ascribed to the addition of ultrafine silica particles, which accelerated the kinetics of hydration and pozzolanic reactions.8-11 This effect was magnified with the incorporation of an accelerating admixture. For example, RF15 and RF30 exhibited final setting times of 130 and 165 minutes shorter than that of NF15 and NF30, respectively. 234

Isothermal calorimetry was conducted on paste samples to complement the trends observed in the setting time test. A rapid rise in the first segment of the heat flow curve may indicate initial setting (end of the dormant period) of the paste, while the peak of the curve indicates its final setting (end of acceleration stage).22,23 For comparison purposes, the heat flow data presented herein are shown over 48 to 80 hours; however, the steady-state stage was developed up to 100 hours. The heat flow and cumulative heat released at 23°C (73°F) for the reference pastes as well as the N and R mixtures are shown in Fig. 2. It can be noted that the heat of hydration curves for the reference samples (without nanosilica) were different from that for the fly ash pastes with nanosilica. The curves for the reference cement pastes were likely dominated by hydration of tricalcium silicate (C3S), similar to the hydration features observed for neat portland cement pastes22,23; however, the nano-modified fly ash pastes (both the N and R mixtures) were typified by a second peak (Fig. 2(b) and 2(c)) with a higher magnitude of heat flow, likely due to the accelerating effect of nanosilica on the reactions. Previous studies attributed the accelerated hydration of C3S to the higher conversion rate of the protective hydrate layer formed close to the C3S surface to a less permeable form due to the abundance of silicate ions24 from nanosilica aggregates (agglomerates of nanosilica in the pore solution, for example due to high pH25 or bridging of silica particles by calcium ions26) and reduction of calcium ions through fast pozzolanic activity (within a few hours).9,27 This led to shortening the induction period and initial setting time of nanomodified fly ash concrete. In addition, it has been postulated that silica aggregates can accelerate the hydration of cement by creating additional surfaces for early precipitation of ACI Materials Journal/March-April 2016

Fig. 2—Isothermal calorimetry curves (normalized heat flow and cumulative heat released) at 23°C (73.4°F): (a) reference mixtures; (b) N mixtures; and (c) R mixtures. (Note: 1 mW/g = 1.548 BTU/(h·lb); 1 J/g = 0.43 BTU/lb.) hydration products and, thus, reducing the final setting time.9,11,27 For the reference pastes, the peak of the heat flow curve was observed at 615 and 675 minutes for the reference pastes comprising 15% and 30% fly ash, respectively (Fig. 2(a)). Comparatively, the hydration of the N nano-modified fly ash pastes was significantly accelerated (Fig. 2(b)) and the length of the dormant period was markedly shortened (by approximately 120 minutes). At a dosage of 6% nanosilica, the first peaks of the hydration curves for NF15 and NF30 were obtained at, respectively, approximately 150 and 180 minutes earlier than that of the corresponding reference pastes. In addition, the total heat evolved over 80  ours was 12 to 18% higher for the N nano-modified fly ash pastes relative to the reference pastes. Again, this catalytic effect of nanosilica was magnified with the incorporation of the accelerating admixture (Fig. 2(c)). For instance, the first ACI Materials Journal/March-April 2016

peaks of the normalized heat flow curves of the RF15, RF22.5, and RF30 mixtures were obtained at, respectively, approximately 175, 155, and 130 minutes earlier than that of the corresponding N mixtures (shifted to the left), and the total cumulative heat of the R mixtures was higher (16 to 21%) than that of the N mixtures. In compliance with the setting time trends that captured the retarding effect of the higher dosage of fly ash, it can be noted that there is a slight delay in the hydration of the mixtures containing 30% fly ash (Fig. 2(b) and 2(c)). The first hydration peaks of mixtures NF30 and RF30 were shifted to the right by approximately 30 and 15 minutes, respectively, relative to that of mixtures NF15 and RF15. Correspondingly, at 80 hours, the total cumulative heat released from the pastes comprising 30% fly ash was lower than that from pastes with 15% fly ash by 18% and 12% for the N and R mixtures, respectively. These results suggest that this retarding effect 235

Table 5—Results of compressive and tensile strengths Mixtures

A

B

NF15

NF22.5

NF30

RF15

RF22.5

RF30

Compressive strength, MPa (psi × 10 ) 3

8h

15 (2.2)

16 (2.3)

—

—

—

10 (1.5)

8 (1.2)

—

1 day

16 (2.3)

18 (2.6)

14 (2.0)

13 (1.9)

10 (1.5)

20 (2.9)

17 (2.5)

13 (1.9)

3 days

21 (3.0)

25 (3.6)

24 (3.5)

20 (2.9)

18 (2.6)

32 (4.6)

30 (4.3)

22 (3.1)

7 days

28 (4.1)

35 (5.1)

31 (4.5)

30 (4.3)

27 (3.9)

38 (5.5)

37 (5.3)

31 (4.5)

28 days

34 (4.9)

38 (5.5)

41 (5.9)

43 (6.2)

36 (5.2)

42 (6.1)

41 (5.9)

38 (5.5)

56 days

35 (5.1)

43 (6.2)

43 (6.2)

45 (6.5)

47 (6.8)

44 (6.4)

45 (6.5)

42 (6.1)

Tensile strength, MPa (psi) 1 day

1.1 (160)

1.8 (260)

1.4 (205)

1.3 (190)

0.9 (130)

2.1 (305)

1.9 (275)

1.2 (175)

3 days

1.9 (275)

2.8 (405)

2.6 (375)

2.7 (390)

1.9 (275)

2.8 (405)

2.8 (405)

2.3 (335)

7 days

2.8 (405)

3.1 (450)

3.8 (550)

3.7 (335)

3.5 (505)

3.8 (550)

3.2 (465)

2.7 (390)

28 days

3.2 (465)

4.1 (595)

5.1 (740)

5.3 (770)

5.7 (825)

4.9 (710)

5.2 (755)

5.5 (800)

was marginal, because it was discounted by the addition of nanosilica, which markedly sped up the kinetics of reactions, as discussed previously. Hence, the addition of nanosilica to the fly ash binders enhanced the hydration level and shortened the hardening time, which affects the early-age strength, as will be discussed in the next section. Strength Table 5 shows the average compressive and splitting tensile strengths of specimens from all mixtures at different ages. The results generally indicated that the commercial products (A and B) gained the highest compressive strength at 8 hours; however, the increase in strength for these products was not significant until 1 day, while the RF15 and RF22.5 specimens achieved higher or comparable compressive strengths of 20 and 17 MPa (2900 and 2465 psi), respectively, at 1 day. In addition, the rate of strength development for products A and B was insignificant up to 3 days. This was statistically supported by the analysis of variance (ANOVA) at a significance level α = 0.05.28 For example, for Product A, ANOVA for the compressive results at 8 hours and 3 days had an F value of 6.71, which is smaller than the critical value (Fcr) of 7.71. This insignificant change in compressive strength up to 3 days may be attributed to the very rapid reactions during the first 8 hours, which might have formed thick hydration shells that discounted the diffusion of water and kinetics of hydration afterward. Comparatively, the rate of strength development of the nano-modified mixtures was significant (an increase of 53 to 80%) up to 3 days. For instance, ANOVA for the compressive strength at 1 and 3 days for the NF15 and RF15 specimens yielded F values of 28.6 and 33.7, respectively, which are larger than the corresponding Fcr of 7.71. The results of the nano-modified mixtures at 28 and 56 days indicated that there was continuous and significant improvement in strength beyond 7 days in comparison to the commercial products. The significant increase in strength of these mixtures with time can be ascribed to the synergistic effects of nanosilica and fly ash, as will be discussed later in the TG results. For specimens from the R mixtures, the average early-age (up to 7 days) strength significantly increased (15 to 43%) 236

in comparison to the N mixtures. This is ascribed to the presence of the accelerating admixture, which sped up the rate of hydration reactions and consequently increased the early-age strength. However, this trend diminished between 7 to 56 days for the R mixtures, which conforms to effect of accelerators on the compressive strength development of concrete at later ages due to the slower diffusion of water through thicker hydration products.22,23 Although increasing the dosage of fly ash in the mixtures led to reducing the compressive strength at early age, the N and R specimens with 30% fly ash (NF30 and RF30) had compressive strength values of 18 and 22 MPa (2610 and 3190 psi) at 3 days, respectively. For the NF30 mixture, the compressive development was significant after 7 days (70% increase between 7 to 56 days) and achieved the highest strength at 56 days (47 MPa [6817 psi]), which is consistent with the well-known effect of Class F fly ash on compressive strength of concrete.6 The early-age (1 to 7 days) results generally indicate that the compressive strength of the N and R nano-modified fly ash mixtures markedly improved with the addition of nanosilica, as no low values were observed for any mixture. These results are consistent with other studies.8,10 Hence, the slow rate of strength development for concrete incorporating Class F fly ash can be controlled by the addition of a small dosage of nanosilica. It appears that nanosilica aggregates effectively contributed to the strength development of the mixtures up to 7 days through high pozzolanic activity,11 filler effect,29,30 and water absorption.30 Moreover, nanosilica can catalyze the reactivity of fly ash at earlyage,8 while the long-term improvement in strength of the mixtures can be ascribed to the continual pozzolanic effect of fly ash with time.6 These mechanisms are discussed in detail later in the TG and microscopy analyses section. The splitting tensile strength of concrete mixtures was determined at different ages, as listed in Table 5. The early-age tensile strength of product A was low, and this product had the lowest tensile strength at 28 days. Comparatively, product B gained higher tensile strength (1.8 MPa [261 psi]) at 1 day; in addition, the increase of tensile strength for product B was significant up to 28 days. Complying with the compressive strength results, all the nano-modified fly ash concrete ACI Materials Journal/March-April 2016

Fig. 3—Bond strength of repair assembly from pulloff test. (Note: 1 MPa = 145 psi; error bars in this figure represent standard error.) mixtures exhibited comparable or higher tensile strength up to 7 days (2.7 to 3.8 MPa [391 to 551 psi]) and 28 days (4.9 to 5.7 MPa [710 to 827 psi]) with a significant rate of increase. The effect of the accelerating admixture on the early-age results of tensile strength was similar to that of the compressive strength. Also, it was observed that the tensile strength slightly decreased with the fly ash content at early age; however, this trend was reversed at 28 days due to the continual reactivity of fly ash, as the N and R specimens with 30% fly ash had the highest tensile strength (Table 5). This behavior may be attributable to the densification of the interfacial transition zone (ITZ) between aggregate and cement paste as a result of the combined effects of nanosilica and fly ash, as shown later in the microscopy analysis section. Bonding Bond failure is a critical cause of deterioration in pavement repairs. Therefore, the pulloff test was used to assess the bond strength of the repair mixtures with substrate concrete; this represents a severe scenario for the assembly, as it is subjected to a direct tension configuration. Also, the pulloff test was used to evaluate the residual bond strength of the mixtures to substrate concrete after combined cyclic environments, which should capture performance risks originating from incompatibility between the repair mixtures and substrate concrete. Therefore, a combined exposure protocol was adopted to replicate consecutive winter and summer seasons, which correlate to in-service conditions. Li et al.31 used a similar technique to evaluate the bonding performance of rapid-setting repair materials subjected to F/T cycles. The average bond strength values before and after the combined exposure are shown in Fig. 3. The average value of pulloff strength from specimens (four cores) produced a coefficient of variation of less than 20% (except for product A). The results showed that the commercial products A and B had bond strength of 1.3 and 1.6 MPa (188 and 232 psi), respectively. The bond strength of these products decreased significantly after the combined exposure by 48 and 31%, respectively, relative to the initial values. The failure of these specimens occurred mainly at the interface between the repair products and substrate concrete (reflecting some level ACI Materials Journal/March-April 2016

of incompatibility) or in the repair products due to their lower tensile capacity (Table 5). Comparatively, the bond strength for the N and R mixtures ranged between 1.9 to 3.3 MPa (275 to 478 psi), and it significantly increased by 23% to 43% after the combined exposure. In addition, the failure in specimens prepared with the N and R mixtures shifted toward the substrate concrete, suggesting that the assembly behaved as an integral system with a high degree of compatibility. Unlike the commercial products, the nano-modified fly ash system offset the deleterious debonding effects induced by cyclic environments and improved the later-age bonding with substrate concrete. This improvement in the bond strength after the combined exposure is thought to be attributed to the continual reactivity of the ternary binder with time of exposure, which might be facilitated by the enhanced moisture level in the repair concrete due to the use of a salt solution rather than fresh water.32 The evolution of the bond strength significantly increased (by 45% to 57%) for the R mixtures in comparison to corresponding specimens from the N mixtures. This may be ascribed to the presence of the accelerating admixture, which sped up the rate of hydration reactions and consequently improved the bond at the interface with substrate concrete since the early stage. In addition, increasing the dosage of fly ash led to increasing the bond strength notably, especially for the R mixtures, as depicted in Fig. 3. This improvement in bonding of the assembly might stem from the chemical interaction between nanosilica and fly ash with the available Ca(OH)2 in the substrate concrete forming secondary calcium-silicate-hydrate (C-S-H) at the bond interface and, thus, enhancing the mechanical interlock between the two layers. The latter argument is substantiated by the efficient reactivity of binder and densified microstructure observed for mixtures comprising higher dosages of fly ash, as shown by the RCPT and thermal and microscopy tests. Penetrability The penetrability of all repair mixtures was evaluated by the RCPT at 28 days, and the results of passing charges and physical penetration depth are listed in Table 6. The commercial product A had a high passing charge value, indicating coarse and continuous pore structure despite 237

achieving a compressive strength of 34 MPa (4930 psi) at this age. According to the classification of ASTM C1202,20 all the nano-modified fly ash concrete mixtures had “very low” penetrability, as their passing charges were below 1000 coulombs, with comparable performance among the N and R mixtures incorporating similar dosages of fly ash. Correspondingly, these mixtures had markedly lower penetration depths (less than 10 mm [0.39 in.]) relative to that of the repair products A (50 mm [1.97 in.]) and B (15 mm [0.59 in.]). The dosage of fly ash in the mixtures had a significant effect on the penetration depth. Considerable reduction of penetrability was achieved for the N and R mixtures containing 30% fly ash, which had comparable compressive strength to that of the commercial repair products at 28 days. ANOVA for the penetration depth results showed that increasing the fly ash content from 15% to 30% in the N and R mixtures had F values 50.3 and 46.2, which are more than the critical value Fcr of 4.1. This trend indicates the densification and discontinuity of the pore structure, which can be explained by the effects of nanosilica and fly ash, as shown later in the thermal and microscopy analyses section, thus improving the durability of the mixtures to the ingress of fluids and, in turn, their projected long-term performance.

Table 6—Rapid chloride penetrability test (RCPT) results Mixture ID

Passing charges, coulombs

Chloride ion penetrability class (ASTM C1202)

A

> 4000

High

50 [0]

B

1443

Low

15 [1.0]

NF15

772

Very low

8 [0.4]

NF22.5

621

Very low

5 [0.7]

NF30

522

Very low

3 [0.3]

RF15

921

Very low

9 [0.7]

RF22.5

644

Very low

6 [0.7]

RF30

602

Very low

5 [0.4]

Note: 1 mm = 0.0394 in.

Average penetration depth, mm [standard error]

Surface scaling The results of surface scaling due to the combined action of deicing salt and F/T cycles (ASTM C672)19 are shown in Fig. 4 and 5. Product A showed a high tendency to surface scaling after approximately 10 cycles. After 50 cycles, products A and B had cumulative mass losses of 2.4 kg/m2 (0.49 lb/ft2) (visual rating of 4 to 5) and 0.5 kg/m2 (0.10 lb/ft2) (visual rating of 1 to 2), receptively. Bureau du normalization du Quebec (BNQ)33 and the Ministry of Transportation, Ontario (MTO)34 stipulate that the failure limits in salt-frost scaling tests are 0.50 and 0.80 kg/m2 [0.10 and 0.16 lb/ft2], respectively. According to these criteria, considering the difference in procedures among the three tests (for example, BNQ and MTO use a less aggressive solution of 3% sodium chloride), products A and B are deemed unacceptable, as they will likely have surface scaling issues in the field. In contrast, all the nano-modified concrete mixtures had limited surface scaling (maximum mass loss of 0.25 kg/m2 [0.05 lb/ft2] [Fig. 4]) and low visual ratings (0 to 1 [Fig. 5]), without a significant difference between the N and corresponding R mixtures. Minor popouts were observed, likely due to the deterioration of some porous aggregates near the surface of concrete. It has been reported that the resistance to salt-frost scaling decreases with increasing the dosage of fly ash in concrete, which is among the key reasons that deter the wider use of higher dosages of fly ash in concrete pavements.5-7 High contents of Class F fly ash in concrete may lead to significant proportions of unbound fly ash particles in the paste, resulting in coarse microstructure and higher tendency to surface scaling.5-7 In the current study, incorporation of 30% fly ash in concrete led to a marginal increase in surface scaling as the binders were modified with nanosilica, indicating improved durability. This was shown by ANOVA for the results of surface scaling that showed a statistically insignificant difference between 15% and 30% fly ash in the N and R mixtures at 50 F/T cycles, as the F values were 3.8 and 9.4, respectively, which are less than the Fcr of 18.51. In addition, when the curing method and time were varied (Table 7), the mass loss results for Mixtures NF30 and RF30 were still below 0.5 kg/m2 (0.10 lb/ft2). The limited surface scaling of the concrete mixtures can be attributed to the incorporation of nanosilica with fly ash, which enhanced the reactivity and

Fig. 4—Mass loss of slabs tested according to ASTM C672. (Note: 1 g/m2 = 2.05 × 10–4 lb/ft2.) 238

ACI Materials Journal/March-April 2016

Table 7—Effect of curing time and method on average mass loss results after 50 freezing-and-thawing cycles Mass loss, g/m2 Curing method

NF15

NF22.5

NF30

RF15

RF22.5

RF30

14 days curing and 14 days in air (ASTM C672)

123

132

189

196

226

247

*

3 days curing and 14 days in air

147

229

398

152

248

410

Curing* the concrete until strength of 20 MPa and 14 days in air

142

221

331

210

363

489

Curing by a chemical compound† and 14 days in air

163

177

236

241

274

376

*

*

Maintained at a temperature of 23 ± 2°C (73.4 ± 35.6°F) and relative humidity of more than 95%.

†

Curing compound met CSA A23.1-14 (Clause 7.7.2.2) specifications.

Note: 1 g/m2 = 2.05 × 10–4 lb/ft2.

Fig. 5—Exemplar visual ratings of slabs after 50 freezing-and-thawing cycles. binding of Class F fly ash, as indicated, for example, by the evolution of compressive and tensile strengths. According to the glue-spall theory,35 which employs fracture mechanics to explain the process of salt-frost scaling by crack propagation into the surface of concrete, the strength of cementitious matrix substantially controls its resistance to scaling. The combination of fly ash with nanosilica led to a dense matrix/ITZ (further densifying with time) and higher tensile capacity, as shown by various tests (for example, bonding, RCPT, BSEM), which might discount the process of crack propagation in the surface of concrete and, thus, improve its resistance to salt-frost scaling. Thermal and microscopy analyses The consumption of portlandite (CH) in the cementitious matrix was determined at different ages to capture the evolution of hydration and pozzolanic reactions, as shown in Fig. 6. At a constant dosage of nanosilica (6%), consumption of CH in the N and R mixtures started at a very early age. For example, at 1 day, the normalized CH contents for NF15 and NF30 relative to the corresponding reference mixtures were less than 1.0 (Fig. 6), which may be linked to a vigorous pozzolanic activity at early age, as observed in the heat of hydration (Fig. 2) and strength tests (Table 5). It has been ACI Materials Journal/March-April 2016

postulated that a very rapid pozzolanic activity is possible as silicate ions from nanosilica aggregates engage with CH forming pozzolanic C-S-H gel, which subsequently precipitate on the surface of silica aggregates, resulting in slower reactivity at later ages.9,27,30 Whether this mechanism is a through-solution process9 or topochemical growth30 is still debatable. Moreover, nanosilica can accelerate the hydration of cement by creating additional surfaces for early precipitation of hydration products.9,11,27 Also, It has been shown that commercial nanosilica sols (originally dispersed to their primary sizes) form small-enough agglomerates to impart a filler effect in the cementitious matrix.29,30 All these factors might have contributed to improving the early-age strength (up to 7 days) of nano-modified fly ash concrete, even for mixtures incorporating 30% fly ash (Table 5). It has been documented that the pozzolanic effect of fly ash in concrete starts at later ages5,22,23 (typically after 28 days; F15 and F30 in Fig. 6); therefore, most standard codes for concrete (for example, CSA A23.1 2014)36 require the properties of fly ash concrete to be assessed at 56 or 91 days. It was reported11 that the pozzolanic action of nanosilica aggregates is completed within 7 days; however, Fig. 6 shows a significant consumption in CH in the N and R mixtures from 7 to 90 days relative to the reference mixtures, especially at 239

Fig. 6—Thermogravimetry results for portlandite content (at temperature range of 400 to 450°C [752 to 842°F]) in nanomodified fly ash concrete mixtures.

Fig. 7—BSEM analysis for a thin section from product A, showing: (a) porous ITZ and coarse microstructure; and (b) associated EDX spectrum of C-S-H in locations indicated in (a). (Note: S.E. is standard error.) higher dosages of fly ash (NF30 and RF30). This pinpoints that the presence of nanosilica catalyzed the reactivity of fly ash in concrete, resulting in an improved level of hydration and an evolution of microstructure. The continual reactivity of the ternary binder up to 90 days is attributed to the pozzolanic activity of fly ash with time, as, for example, noted in the results of later-age strength (Table 5) and bonding after the combined exposure (Fig. 3). This explains the significant densification and refinement of the pore structure at 28 days in the nano-modified fly ash concrete mixtures, which had limited penetration depth (average of 6 mm [0.24 in.]). BSEM was conducted on thin sections to complement the trends observed in the mechanical, durability, and TG tests. In comparison to the commercial product A, all the nano-modified fly ash mixtures had a significant degree of refinement and densification in the hydrated paste and ITZ at 28 days. Product A (Fig. 7) showed coarse microstructure in addition to microcracks in the paste and ITZ; the EDX analysis for C-S-H in the ITZ showed a high calcium-silicate ratio (C/S) (average of 2.0). These features reflect an insufficient level of hydration conforming to the inferior performance observed for this product in the mechanical and durability tests. On the contrary, homogenous and dense matrix in various specimens from the N and R mixtures with low (15%) and high 240

(30%) dosages of fly ash were observed. For instance, Fig. 8 shows dense microstructure and refined ITZ at 28 days of a specimen from NF30 owing to the synergistic effects of nanosilica and fly ash, as described previously. EDX analysis for C-S-H in the ITZ in this specimen showed that the average C/S was 1.05, indicating an efficient pozzolanic activity, and densification of ITZ with secondary C-S-H. It was reported that the C/S of secondary C-S-H from pozzolanic reactions is lower than that of conventional C-S-H produced from cement hydration reactions, the former has a ratio of approximately 1.1, whereas the latter has a ratio of approximately 1.7.37 In addition, Kong et al.30 stated that when small agglomerates of nanosilica form, water absorption into their ultra-high nano-porosity can reduce the waterbinder ratio (w/b) in the paste, thus improving the microstructure of the matrix. CONCLUSIONS Considering the materials, mixture designs, and testing methods implemented in this study, the following conclusions can be drawn: 1. This study indicates that the high early-strength of rapid-setting materials is an insufficient criterion to consider a product acceptable as a repair material for concrete paveACI Materials Journal/March-April 2016

Fig. 8—BSEM analysis for a thin section from NF30, showing: (a) refined ITZ and dense microstructure; and (b) associated EDX spectrum of C-S-H in locations indicated in (a). (Note: S.E. is standard error.) ments. Except for early-age strength, the commercial product A had adverse performance in many aspects such as placement/ finishability and salt-frost scaling. Comparatively, product B had better performance in early-age strength and resistance to the ingress of fluids and salt-frost scaling. 2. The normal and rapid nano-modified fly ash concrete mixtures developed herein had ample hardening times, without excessive delay, which improves the flexibility and quality of the repair process with suitability for different repair applications (for example, multiple/large patch areas with less critical opening time to traffic). 3. The incorporation of 6% nanosilica in concrete with up to 30% fly ash significantly shortened the dormant period and accelerated the rate of hydration reactions, which discounted some of the retarding effect of Class F fly ash on the rate of hardening of concrete. 4. The synergistic effects of nanosilica and fly ash in both the N and R mixtures improved the early-age and long-term compressive, tensile, and bond (even after the combined exposure) strengths of concrete, which indicate that the inherently slower rate of strength and microstructural development of concrete containing Class F fly ash can be controlled by the addition of small dosages of nanosilica. 5. The combination of fly ash with nanosilica led to a dense pore structure (as shown by the low penetration depth in the RCPT) and improved tensile capacity (as reflected by the limited surface scaling of concrete), which suggest that the addition of nanosilica with fly ash enhanced its binding in the matrix. 6. In addition to the effects of nanosilica on improving the hydration and pore structure characteristics by multiple mechanisms, TG results showed that addition of 6% nanosilica to concrete incorporating up to 30% fly ash catalyzed the reactivity of Class F fly ash, resulting in an improved level of hydration and an evolution of microstructure with age. This explains the significant densification/refinement in the hydrated paste and ITZ, as shown by BSEM at 28 days, and the corresponding improved mechanical and durability performance of the nano-modified fly ash concrete at different ages. The overall results indicate that nano-modified fly ash concrete can achieve balanced early-age and long-term performance. Hence, it presents a viable option for a suite ACI Materials Journal/March-April 2016

of repair applications in concrete pavements, with an anticipated measurable impact on reducing life cycle costs. Yet, its field performance needs to be documented, which is recommended for future work. AUTHOR BIOS

ACI member A. Ghazy is a PhD Candidate at the University of Manitoba, Winnipeg, MB, Canada. He received his BSc and MSc from Alexandria University, Egypt. His research interests include durability of concrete pavements. ACI member M. T. Bassuoni is an Associate Professor in the Department of Civil Engineering at the University of Manitoba. He is a member of ACI Committees 201, Durability of Concrete; 236, Material Science of Concrete; 237, Self-Consolidating Concrete; and 241, Nanotechnology of Concrete. His research interests include cementitious materials and durability of concrete. A. Shalaby is a Professor in the Department of Civil Engineering at the University of Manitoba. His research interests include pavement engineering and infrastructure management.

ACKNOWLEDGMENTS

The authors highly appreciate the financial support from Natural Sciences and Engineering Research Council of Canada, University of Manitoba Graduate Fellowship, and City of Winnipeg. The new IKO Construction Materials Testing Facility at the University of Manitoba in which these experiments were conducted has been instrumental to this research.

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Demis, Sotiris University of Patras Patra, Greece Den Uijl, Joop Delft University of Technology Delft, the Netherlands Deng, Yaohua Iowa State University Ames, IA, United States Detwiler, Rachel Braun Intertec Corp. Minneapolis, MN, United States Dheilly, Rose-Marie Laboratoire des Technologies Innovantes Dury, France G. Dhinakaran Sastra University Thanjavur, India Dhole, Rajaram St. John’s, NL, Canada Dhonde, Hemant University of Houston Houston, TX, United States Di Ludovico, Marco University of Naples Federico II Naples, Italy Diao, Bo Beihang University Beijing, China Diaz Loya, Eleazar Louisiana Tech University Ruston, LA, United States Dinev, Dobromir UACEG Sofia, Bulgaria Diniz, Sofia Maria Universidade Federal de Minas Gerais Be lo Horizonte, Brazil Dogan, Unal Istanbul Technical University Istanbul, Turkey Doh, Jeung-Hwan Griffith University Gold Coast, Queensland, Australia Dolan, Charles University of Wyoming Laramie, WY, United States Dongell, Jonathan Pebble Technologies Scottsdale, AZ, United States

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Du, Hongjian National University of Singapore Singapore Du, Jinsheng Beijing Jiao Tong University Beijing, China Du, Lianxiang The University of Alabama at Birmingham Birmingham, AL, United States Dutta, Anjan Indian Institute of Technology Guwahati Guwahati, Assam, India Duvallet, Tristana University of Kentucky Lexington, KY, United States El-Ariss, Bilal United Arab Emirates University Al Ain, United Arab Emirates El-Dash, Karim College of Technological Studies Kuwait El-Dieb, Amr Ain Shams University Abbasia, Cairo, Egypt El-Hawary, Moetaz Kuwait Institute for Scientific Research Safat, Kuwait El-Maaddawy, Tamer United Arab Emirates University Al-Ain, Abu Dhabi, United Arab Emirates El-Metwally, Salah University of Hawaii at Manoa Honolulu, HI, United States El-Refaie, Sameh El-Gama City, Mataria, Cairo, Egypt El-Salakawy, Ehab University of Manitoba Winnipeg, MB, Canada El-Sayed, Ahmed University of Sherbrooke Sherbrooke, QC, Canada Elahi, Ayub Taxilla, Pakistan Elamin, Anwar University of Nyala Nyala, Sudan

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Elbahar, Mohamed Ken Okamoto & Associates - Stuctural Engineers Rancho Santa Margarita, CA, United States

Ferguson, Bruce University of Georgia Athens, GA, United States

Eldarwish, Aly Alexandria, Egypt

Fernandez Montes, David Madrid, Spain

Elfgren, Lennart Luleå University of Technology Luleå, Sweden

Fernández Ruiz, Miguel Ecole Polytechnique Federale de Lausanne Lausanne, Vaud, Switzerland

Elmasry, Mohamed Arab Academy for Science and Technology and Maritime Transport Alexandria, Egypt

Ferrara, Liberato Politecnico di Milano Milan, Italy

Elsayed, Tarek Cairo, Egypt

Ferraris, Chiara NIST Gaithersburg, MD, United States

Eltahawy, Reham Ain Shams University Cairo, Egypt

Ferron, Raissa University of Texas at Austin Austin, TX, United States

Erdogan, Sinan Middle East Technical University Ankara, Turkey

Folino, Paula University of Buenos Aires Buenos Aires, Argentina

Erdogan, Yasin Hatay, Iskenderun, Turkey

Foraboschi, Paolo Universita IUAV di Venezia Venice, Italy

Ergün, Ali Technical Education Faculty Afyonkarahısar, Turkey Esmaeily, Asad Kansas State University Manhattan, KS, United States Esperanza, Menendez IETCC-CSIC Madrid, Spain Fantilli, Alessandro Politecnico di Torino Torino, Italy Fardis, Michael Patras, Greece Farghaly, Ahmed University of Sherbrooke Sherbrooke, QC, Canada Faria, Duarte Caparica-Lisbon, Portugal Farrokhi, Farhang Zanjan, Islamic Republic of Iran Farzam, Masood Tabriz, Islamic Republic of Iran Fathi, Hamoon Sanandaj Branch, Islamic Azad University Sanandaj, Kurdistan, Islamic Republic of Iran Fattuhi, Nijad Birmingham, UK Feldman, Lisa University of Saskatchewan Saskatoon, SK, Canada Fenollera, Maria Universidade de Vigo Vigo, Spain

ACI Materials Journal/March-April 2016

Forth, John University of Leeds Leeds, UK Fouad, Fouad University of Alabama at Birmingham Birmingham, AL, United States Francüois, Buyle-Bodin University of Lille Villeneuve d’Ascq, France Frosch, Robert Purdue University West Lafayette, IN, United States Fuchs, Werner University of Stuttgart Stuttgart, Germany Fuentes, Jose Maria Polytechnic University of Madrid Madrid, Spain Fujikake, Kazunori National Defense Academy Yokosuka Kanagawa, Japan Gabrijel, Ivan University of Zagreb Zagreb, Croatia Gajdosova, Katarina Bratislava, Slovakia Ganesan, N. National Institute of Technology Calicut, India Gao, XiangLing Tongji University Shanghai, China

249

REVIEWERS IN 2015 Garcez, Estela Universidade Federal de Pelotas Pelotas, RS, Brazil

Guo, Zixiong Huaqiao University Quanzhou, Fujian, China

Gayed, Ramez University of Calgary Calgary, AB, Canada

Gupta, Pramod I.I.T. Roorkee Roorkee, Uttarakhand, India

Gergely, Janos UNC Charlotte Charlotte, NC, United States

Gupta, Rajiv BITS Pilani Pilani, Rajasthan, India

Gesoglu, Mehmet Gaziantep University Gaziantep, Turkey

Gupta, Rishi Vancouver, BC, Canada

Ghezal, Aïcha Ecole de Technologie de Montreal Montreal, QC, Canada Giaccio, Craig AECOM Melbourne, Victoria, Australia Girgin, Canan Yildiz Technical University Istanbul, Turkey Goel, Rajeev CSIR-Central Road Research Institute Delhi, India Gongxun, Wang Hunan University of Science and Technology Xiangtan, China González, Javier University of Basque Country Bilbao, Basque Country, Spain González, María Polytechnic University of Madrid Madrid, Spain

Gursel, Aysegul University of California, Berkeley Berkeley, CA, United States Haach, Vladimir University of São Paulo São Carlos, São Paulo, Brazil Habbaba, Ahmad Technische Universität München Garching, Germany Haber, Zachary PSI McLean, VA, United States Haddadin, Laith United Nations New York, NY, United States Hadi, Muhammad University of Wollongong Wollongong, New South Wales, Australia Hadje-Ghaffari, Hossain John A. Martin & Assoc. Los Angeles, CA, United States

Gonzalez-Valle, Enrique Madrid, Spain

Hagenberger, Michael The Ohio State University Columbus, OH, United States

Grandić, Davor University of Rijeka Rijeka, Croatia

Hager, Angela City and County of Denver Denver, CO, United States

Gribniak, Viktor Vilnius Gediminas Technical University Vilnius, Lithuania

Haggag, Hesham Cairo, Egypt

Gross, Shawn Villanova University Villanova, PA, United States Guadagnini, Maurizio The University of Sheffield Sheffield, UK Güneyisi, Erhan Gaziantep University Gaziantep, Turkey Guo, Honglei Wuhan Polytechnic University Wu Han City, Hu Bei Province, China Guo, Liping Southeast University Nanjing, Jiangsu Province, China

250

Hamid, Roszilah Universiti Kebangsaan Malaysia Bangi, Selangor, Malaysia Haneefa Kolakkadan, Mohammad SSN College of Engineering Kalavakkam, Tamilnadu, India Hao, Yifei Curtin University Bentley, Western Australia, Australia Harajli, Mohamed American University of Beirut Beirut, Lebanon Harbec, David Université de Sherbrooke Sherbrooke, QC, Canada

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Hariri-Ardebili, Mohammad Amin University of Colorado Boulder, CO, United States

Hrynyk, Trevor University of Texas at Austin Austin, TX, United States

Harmon, Thomas Washington University in St. Louis Saint Louis, MO, United States

Hu, Jiong University of Nebraska – Lincoln Lincoln, NE, United States

Harries, Kent University of Pittsburgh Pittsburgh, PA, United States

Hu, Nan Tsinghua University Beijing, China

Hashemi, Shervin Seoul National University Seoul, Republic of Korea

Huang, Chang-Wei Chung Yuan Christian University Chung Li, Taiwan, China

Hassan, Assem Toronto, ON, Canada

Huang, Chung-Ho National Taipei University of Technology Taipei, Taiwan, China

Hassan, Maan University of Technology Baghdad, Iraq Hassan, Mohamed University of Sherbrooke Sherbrooke, QC, Canada Hawileh, Rami American University of Sharjah Sharjah, United Arab Emirates He, Xiaobing Chongqing Jiaotong University Chongqing, China He, Zhiqi Southeast University Nanjing, Jiangsu, China Helene, Paulo Universidade de São Paulo São Paulo, São Paulo, Brazil Helmy, Huda Applied Science International Durham, NC, United States Hemalatha, T. CSIR-Structural Engineering Research Centre Chennai, Tamil Nadu, India Henry, Richard University of Auckland Auckland, New Zealand Ho, Johnny The University of Hong Kong Hong Kong, China Hong, Sung-Gul Seoul National University Seoul, Republic of Korea Hossain, Khandaker Ryerson University Toronto, ON, Canada Hossain, Tanvir Louisiana State University Houston, TX, United States Hover, Kenneth C. Cornell University Ithaca, NY, United States

ACI Materials Journal/March-April 2016

Huang, Jianwei Southern Illinois University Edwardsville Edwardsville, IL, United States Huang, Qindan The University of Akron Akron, OH, United States Hung, Chung-Chan University of Michigan Ann Arbor, MI, United States Hung, Mengfeng De Lin Institute of Technology New Taipei City, Taiwan, China Husain, Mohamed Zagazig University Zagazig, Egypt Hwang, Shyh-Jiann National Taiwan University Taipei, Taiwan, China Ibrahim, Amer Baquba, Iraq Ibrahim, Hisham Buckland and Taylor ltd North Vancouver, BC, Canada Ichinose, Toshikatsu Nagoya Institute of Technology Nagoya, Japan Ilki, Alper Istanbul Technical University Istanbul, Turkey Ince, Ragip Firat University Engineering Faculty Elazig, Turkey Irassar, Edgardo Dep. Ingenieria Civil - UNCPBA Olavarria, Buenos Aires, Argentina Jain, Mohit Nirma University Ahmedabad, Gujarat, India Jain, Shashank Delhi Technological University (DTU) New Delhi, India

251

REVIEWERS IN 2015 Jamkar, Sanjay Govt. College of Engineering, Aurangabad Aurangabad, Maharashtra, India

Kang, Thomas Seoul National University Seoul, Republic of Korea

Jang, Seung Yup Korea Railroad Research Institute Uiwang, Gyongggi-do, Republic of Korea

Kankam, Charles Kwame Nkrumah University of Science & Technology Kumasi, Ghana

Janotka, Ivan Building Testing and Research Institute Bratislava, Slovakia

Kansara, Kunal Mouchel Infrastructure Services Bristol, UK

Jansen, Daniel California Polytechnic State University San Luis Obispo, CA, United States

Kantarao, Velidandi Central Road Research Institute New Delhi, Delhi, India

Jarallah, Husain The University of Mustansiriyah Baghdad, Iraq

Kanwar, Varinder Chitkara University Barotiwala, Himachal Pradesh, India

Jeng, Chyuan-Hwan National Chi Nan University-Taiwan Puli/Nantou, Taiwan, China

Karayannis, Christos Democritus University of Thrace Xanthi, Greece

Jeon, Se-Jin Ajou University Suwon-si, Gyeonggi-do, Republic of Korea

Kashipurad, K. B. Prakash Government Engineering College Haveri, Karnataka, India

Jiang, Hai-Jun State Key Laboratory of Coal Resources and Safe Mining  (CUMT) Xuzhou, Jiangsu, China

Katz, Amnon Technion-Israel Institute of Technology Haifa, Israel

Jiang, Hua University of Georgia Athens, GA, United States Jiang, Jiabiao W. R. Grace (Singapore) Pte Ltd Singapore Jin, Ruoyu University of Nottingham Ningbo China Ningbo, Zhejiang, China Jozić, Dražan Split, Croatia Juvandes, Luis Universidade do Porto: FEUP Porto, Portugal

Kawamura, Mitsunori Kanazawa, Ishikawa, Japan Kazemi, Mohammad Sharif University of Technology Tehran, Islamic Republic of Iran Kazemi, Sadegh University of Alberta Edmonton, AB, Canada Kenai, Said Université de Blida Blida, Algeria Kenel, Albin Rapperswil University of Applied Sciences HSR Horw, Switzerland

Kabashi, Naser Prishtine Kosove, Albania

Kevern, John University of Missouri – Kansas City Kansas City, MO, United States

Kaklauskas, Gintaris Vilnius Gediminas Technical University Vilnius, Lithuania

Khan, Mohammad King Saud University Riyadh, Saudi Arabia

Kam, Weng Yuen University of Canterbury Christchurch, Canterbury, New Zealand

Khan, Sadaqat Universiti Teknologi PETRONAS Tronoh, Perak, Malaysia

Kamanli, Mehmet Selcuk University Konya, Turkey

Khennane, Amar UNSW @ ADFA Canberra, Australian Capital Territory, Australia

Kampmann, Raphael Florida State University Tallahassee, FL, United States

Kheyroddin, Ali University of Texas at Arlington Arlington, TX, United States

Kanagaraj, Ramadevi Kumaraguru College of Technology Coimbatore, Tamilnadu, India

Kianoush, M. Reza Ryerson University Toronto, ON, Canada

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ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Kilic, Sami Bogazici University Istanbul, Turkey

Kumar, Ratnesh Visvesaraya National Institute of Technology Nagpur Nagpur, Maharashtra, India

Kim, Eunhye Colorado School of Mines Golden, CO, United States

Kumar, Vinod Steel Authority of India Limited Ranchi, Jharkhand, India

Kim, Jang Hoon Ajou University Suwon, Republic of Korea

Kunnath, Sashi University of California, Davis Davis, CA, United States

Kim, JunHee Korea Institute of Construction Technology Goyang-si, Gyeonggi-do, Republic of Korea

Kupwade-Patil, Kunal Massachusetts Institute of Technology Cambridge, MA, United States

Kim, Kil-Hee Kongju National University Kongju, Chungnam, Republic of Korea

Kusbiantoro, Andri Universiti Malaysia Pahang Gambang, Pahang, Malaysia

Kim, Sang Heon Konkuk University Seoul, Republic of Korea

Kwan, Albert The University of Hong Kong Hong Kong, China

Kim, Yail Jimmy University of Colorado Denver Denver, CO, United States Kim, Yong Jic Daewoo E&C Suwon, Gyeonggi, Republic of Korea

Kwan, Wai Hoe Universiti Sains Malaysia Gelugor, Penang, Malaysia

Kim, Young Hoon Oregon State University Corvallis, OR, United States

Labibzadeh, Mojtaba Ahvaz, the Islamic Republic of Iran

Kishi, Norimitsu Muroran Institute of Technology Muroran, Japan Kisicek, Tomislav Zagreb, Croatia Kitayama, Kazuhiro Tokyo Metropolitan University Tokyo, Japan Kode, Venkata Ramesh Gitam University Visakhapatnam, Andhra Pradesh, India Koenders, Eddie A. B. TU Darmstadt Darmstadt, Hessen, Germany Kohler, Erwin Dynatest Davis, CA, United States Kotsovos, Gerasimos National Technical University of Athens Athens, Greece Kotsovos, Michael Athens, Greece Koyama, Tomoyuki Fukuoka, Japan Krem, Slamah University of Waterloo Waterloo, ON, Canada Kumar, Rakesh Central Road Research Institute Delhi, India

ACI Materials Journal/March-April 2016

La Tegola, Antonio University of Lecce Lecce, Italy

Lagaros, Nikos Institute of Structural Analysis & Seismic Research, National Technical University of Athens Athens, Greece Lai, James La Canada, CA, United States Lai, Jianzhong Nanjing University of Science and Technology Nanjing, Jiangsu, China Lamanna, Anthony New Orleans, LA, United States Lampropoulos, Andreas University of Patras Patras, Greece Laskar, Aminul National Institute of Technology Silchar, Assam, India Laterza, Michelangelo University of Basilicata Potenza, Italy Lau, Teck University of Nottingham Semenyih, Selangor, Malaysia Law, David RMIT University Melbourne, Victoria, Australia Lawler, John Wiss, Janney, Elstner Associates, Inc. Northbrook, IL, United States

253

REVIEWERS IN 2015 Lee, Chadon Chung-Ang University Ansung, Kyungki-Do, Republic of Korea

Li, Zhu Taiyuan University of Technology Taiyuan, China

Lee, Chi King Nanyang Technological University Singapore

Lignola, Gian Piero University of Naples Federico II Naples, Italy

Lee, Deuck Hang University of Seoul Seoul, Republic of Korea

Lima, Maria Cristina Federal University of Uberlândia Uberlandia MG, Minas Gerais, Brazil

Lee, Douglas Douglas D. Lee and Associates Fort Worth, TX, United States

Lin, Wei-Ting Ilan, Taiwan, China

Lee, Heui Hwang Arup San Francisco, CA, United States Lee, Hung-Jen National Yunlin University of Science and Technology Douliu, Yunlin, Taiwan, China Lee, Jae-Man Lotte Engineering and Construction Seoul, Republic of Korea Lee, Nam Ho SNC-Lavalin Nuclear Oakville, ON, Canada Lee, Seong-Cheol KEPCO International Graduate School (KINGS) Ulsan, Republic of Korea Lee, Yoon-Si Bradley University Peoria, IL, United States Lee, Young Hak Seoul, Republic of Korea Leone, Marianovella University of Salento Lecce, Italy Lequesne, Remy University of Kansas Lawrence, KS, United States Li, Fumin China University of Mining and Technology Xuzhou, Jiangsu, China Li, Jiabin Graz University of Technology Graz, Austria Li, Qingbin Tsinghua University Beijing, China Li, Qixuan Xi’an, China Li, Shuguang China Institute of Water Resources and Hydropower Research Beijing, China Li, Wei Wenzhou University Wenzhou, Zhejiang, China Li, Xinghe University of New Hampshire Durham, NH, United States

254

Lin, Zhibin Fargo, ND, United States Liu, Jiepeng Chongqing University Chongqing, China Liu, Jun Beijing, China Liu, Junshan Sargent Lundy LLC Chicago, IL, United States Liu, Shuhua Wuhan University Wuhan, HuBei, China Liu, Xuejian University of Texas at Arlington Arlington, TX, United States Liu, Yanbo Florida Atlantic University Dania Beach, FL, United States Liu, Ze China University of Mining & Technology, Beijing Beijing, China Liu, Zhao Southeast University Nanjing, Jiangsu, China Lizarazo Marriaga, Juan Coventry University Coventry, UK Lodi, Sarosh NED University Karachi, Pakistan Long, Adrian Queens University Belfast, UK Long, Xu Nanyang Technological University Singapore Loo, Yew-Chaye Gold Coast, Australia Lopes, Anne Furnas Centrais Electricas SA Aparecida de Goiania, Goias, Brazil Lopes, Sergio University of Coimbra Coimbra, Portugal

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

López-Almansa, Francisco Technical University of Catalonia Barcelona, Spain

Manso, Juan University of Burgos Burgos, Castilla - León, Spain

Lotfy, Abdurrahmaan Lafarge Canada Inc. Toronto, ON, Canada

Manzur, Tanvir Bangladesh University of Engineering & Technology Dhaka, Bangladesh

Lou, Lei S and R Engineers, P.C. Hamilton, NJ, United States

Marar, Khaled Eastern Mediterranean University Gazimagusa, Turkey

Lounis, Zoubir National Research Council Ottawa, ON, Canada

Mari, Antonio Universitat Politecnica de Catalunya Barcelona, Spain

Lushnikova, Nataliya National University of Water Management and Nature  Resources Use Rivne, Ukraine

Markovic, Ivan DSP Ingenieure & Planer AG Greifensee, Switzerland

Ma, Zhongguo University of Tennessee Knoxville, TN, United States Maage, Magne Skanska Norge AS Trondheim, Norway Machida, Atsuhiko Saitama University Saitama, Japan

Martinelli, Enzo University of Salerno Fisciano, Italy Martin-Perez, Beatriz University of Ottawa Ottawa, ON, Canada Maslehuddin, Mohammed King Fahd University of Petroleum and Minerals Dhahran, Saudi Arabia

Macht, Jürgen Kirchdorf, Austria

Matsagar, Vasant Lawrence Technological University Southfield, MI, United States

Madani, Hesam Kerman Graduate University of Technology Kerman, Islamic Republic of Iran

Matta, Fabio University of South Carolina Columbia, SC, United States

Maggenti, Ric Caltrans Sacrmento, CA, United States

Maximos, Hany Pharos University in Alexandria Alexandria, Egypt

Magliulo, Gennaro University of Naples Federico II Naples, Italy

Mazzotti, Claudio University of Bologna Bologna, Italy

Maguire, Marc Utah State University Paradise, UT, United States

Mbessa, Michel University of Yaoundé I - ENSP Yaoundé, Center, Cameroon

Mahdy, Mohamed Masoura University Mansoura, Dakhlia, Egypt

Meddah, Mohammed Seddik Kingston University London Kingston, Surrey, London, UK

Mahfouz, Ibrahim Cairo, Egypt

Megally, Sami PBS&J San Diego, CA, United States

Mahrenholtz, Christoph Berlin, Germany Makul, Natt Phranakhon Rajabhat University Bangkok, Thailand Mander, John Texas A&M University College Station, TX, United States Manfredi, Gaetano University of Naples Naples, Italy

ACI Materials Journal/March-April 2016

Mehanny, Sameh Cairo University Cairo, Egypt Mejia, Luis Gonzalo LGM & Cia Medellín, Colombia Melo, José University of Porto Porto, Portugal Meng, Tao Institution of Building Materials Hangzhou, Zhejiang, China

255

REVIEWERS IN 2015 Mermerdaş, Kasım Hasan Kalyoncu University Gaziantep, Turkey

Motaref, Sarira University of Connecticut Storrs, CT, United States

Mezhov, Alexander Moscow State University of Civil Engineering Moscow, Russian Federation

Moyo, Pilate University of Cape Town Cape Town, South Africa

Micelli, Francesco University of Salento Lecce, Italy

Mubin, Sajjad University of Engineering and Technology Lahore, Punjab, Pakistan

Michael, Antonis University of Florida Gainesville, FL, United States

Muciaccia, Giovanni Politecnico di Milano Milan, Italy

Milestone, Neil Callaghan Innovation Lower Hutt, New Zealand

Mukai, David University of Wyoming Laramie, WY, United States

Mishra, Laxmi MNNIT Allahabad, UP, India

Mukherjee, Abhijit Curtin University Bentley, Western Australia, Australia

Misra, Sudhir IIT Kanpur Kanpur, India

Mulaveesala, Ravibabu Indian Institute of Technology Ropar Rupnagar, India

Mo, Xiangyin Nanjing Normal University Nanjing, Jiangsu, China

Mullapudi, Taraka Ravi MMI Engineering Houston, TX, United States

Mohamed, Ashraf Alexandria University Alexandria, Egypt

Muñoz, Jose Federal Highway Administration McLean, VA, United States

Mohammadyan Yasouj, Seyed Esmaeil UTM University Johor, Malaysia

Murty, Devalraju Andhra University Visakhapatnam, AP, India

Montes, Carlos Louisiana Tech University Ruston, LA, United States

Muttoni, Aurelio Swiss Federal Institute of Technology Lausanne, Switzerland

Moradian, Masoud Oklahoma State University Stillwater, OK, United States

Na, Okpin Korea Railroad Research Institute Uiwang-Si, Gyeonggi-do, Republic of Korea

Moreno, Carlos Instituto Politécnico de Bragança Bragança, Portugal

Nadeau, Joseph Duke University Durham, NC, United States

Moreno, Eric Universidad Autónoma de Yucatan Merida, Yucatan, Mexico

Nafie, Amr Cairo, Egypt

Moretti, Marina University of Thessaly Athens, Greece Morley, Christopher Cambridge University Cambridge, UK Moser, Robert U.S. Army Engineer Research and Development Center Vicksburg, MS, United States Mostafaei, Hossein University of Toronto Toronto, ON, Canada Mostofinejad, Davood Isfahan University of Technology Isfahan, Islamic Republic of Iran

256

Naganathan, Sivakumar Universiti Tenaga Nasional Kajang, Selangor, Malaysia Naish, David California State University, Fullerton Fullerton, CA, United States Najim, Khalid University of Anbar Ramadi, Anbar, Iraq Najimi, Meysam University of Nevada, Las Vegas Las Vegas, NV, United States Nakamura, Hikaru Nagoya University Nagoya, Japan

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Nam, Boo Hyun University of Central Florida Orlando, FL, United States

Orr, John University of Bath Bath, UK

Narayanan, Pannirselvam VIT University Vellore, Tamilnadu, India

Orta, Luis ITESM Zapopan, Jalisco, Mexico

Narayanasamy, Rajeswari Universidad Juarez del Estado de Durango Gomez Palacio, Durango, Mexico

Ortiz-Lozano, Jose Autonomous University of Aguascalientes Aguascalientes, Mexico

Negrutiu, Camelia Technical University of Cluj Napoca Cluj Napoca, Cluj, Romania

Osifala, Kehinde Somolu, Lagos, Nigeria

Neves, Luís University of Coimbra Coimbra, Portugal Ng, Ivan Drainage Services Department Hong Kong, China Ng, Pui Lam The University of Hong Kong Hong Kong, China Nguyen, Hai Huntington, WV, United States Nili, Mahmoud Bu-Ali Sina University Hamedan, Islamic Republic of Iran Nimityongskul, Pichai Asian Institute of Technology Pathumtahni, Thailand Nokken, Michelle Concordia University Montreal, QC, Canada Noor, Munaz Bangladesh University of Engineering and Technology Dhaka, Bangladesh Noshiravani, Talayeh EPFL Lausanne, Switzerland Oh, Hongseob Jinju National University Jinju, Kyeongnam, Republic of Korea Okelo, Roman Dallas, TX, United States Oliva, Michael University of Wisconsin Madison, WI, United States Ombres, Luciano University of Calabria Cosenza, Italy Omran, Ahmed University of Sherbrooke Sherbrooke, QC, Canada Orakcal, Kutay Bogazici University Istanbul, Bebek, Turkey

ACI Materials Journal/March-April 2016

Otsuki, Nobuaki Tokyo Institute of Technology Tokyo, Japan Ou, Yu-Chen National Taiwan University of Science and Technology Taipei, Taiwan, China Oyamada, Tetsuya Iwate University Morioka, Iwate, Japan Ozbay, Erdogan Iskenderun, Hatay, Turkey Ozbolt, Josko Stuttgart, Germany Ozturan, Turan Bogazici University Istanbul, Turkey Pacheco, Alexandre Universidade Federal do Rio Grande do Sul (UFRGS) Porto Alegre, RS, Brazil Palaniraj, Saravanakumar Sastra University Thanjavur, Tamilnadu, India Palazzo, Gustavo National Technological University Ciudad Mendoza, Mendoza, Argentina Palermo, Dan University of Ottawa Ottawa, ON, Canada Palieraki, Vasiliki National Technical University of Athens Athens, Zografou, Greece Palmisano, Fabrizio Politecnico di Bari Bari, Italy Pan, Zuanfeng Tongji University Shanghai, China Pandit, Poorna National Institute of Technology Karnataka Mangalore, Karnataka, India Pang, Xueyu Halliburton Houston, TX, United States Pantazopoulou, Stavroula Demokritus University of Thrace Xanthi, Greece

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REVIEWERS IN 2015 Pantelides, Chris Salt Lake City, UT, United States Pape, Torill University of Newcastle Callaghan, New South Wales, Australia

Pisani, Marco Politecnico di Milano Milan, Italy Pocesta, Ylli Debar, the Former Yugoslav Republic of Macedonia

Parameswaran, Lakshmy CSIR-Central Road Research Institute New Delhi, Delhi, India

Popovics, John University of Illinois Urbana, IL, United States

Parghi, Anant The University of British Columbia Kelowna, BC, Canada

Potter, William Florida Department of Transportation Tallahassee, FL, United States

Park, Ki-bong Chuncheon-Si, the Republic of Korea

Pourazin, Khashaiar Pars AB Tadbir Consulting Engineers Co. Tehran, Islamic Republic of Iran

Parsekian, Guilherme Federal University of São Carlos São Carlos, São Paulo, Brazil Patil, Sanjay Caledonian College of Engineering Seeb, Oman Pauletta, Margherita University of Udine Tavagnacco, Udine, Italy Peixoto, Ricardo Andre Federal University of Ouro Preto Ouro Preto, Minas Gerais, Brazil Pellegrino, Carlo University of Padova Padova, Italy Peng, Cao Harbin Institute of Technology Harbin, Heilongjiang, China Pereira, Eduardo University of Minho Guimaraes, Portugal Pereyra, María Instituto de Estructuras y Transporte Montevideo, Uruguay Perez Caldentey, Alejandro Universidad Politécnica de Madrid Madrid, Spain Perez Ruiz, Diego Pontificia Universidad Javeriana Sede Cali Cali, Valle Del Cauca, Colombia Persson, Bertil Bara, Sweden Petrone, Floriana University of California, Davis Davis, CA, United States Phillippi, Donald Kansas State University Manhattan, KS, United States Piccinin, Roberto Hilti, Inc. Tulsa, OK, United States Pilakoutas, Kypros The University of Sheffield Sheffield, UK

258

Prakash, M. N. J.N.N. College of Engineering Shimoga, Karnataka, India Prashanth, P. SJCE Mysore, Karnataka, India Prasittisopin, Lapyote Oregon State University Corvallis, OR, United States Premalatha, P. V. CARE School of Engineering Tiruchirappalli, India Prusinski, Jan Slag Cement Association Sugar Land, TX, United States Puertas, F. Eduardo Torroja Institute Madrid, Spain Puthenpurayil Thankappan, Santhosh Granite Construction Company Abu Dhabi, United Arab Emirates Putra Jaya, Ramadhansyah Universiti Teknologi Malaysia Skudai, Johor Bahru, Malaysia Qasrawi, Hisham The Hashemite University Zarqa, Jordan Qi, Guihai Guiyang City, Guizhou Province, China Qian, Kai NTU Singapore Qu, Jili University of Shanghai for Science and Technology Shanghai, China Quiroga, Pedro Escuela Colombiana de Ingenieria Bogota, Colombia Rafi, Muhammad NED University of Engineering and Technology Karachi, Sindh, Pakistan

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Ragheb, Wael Alexandria University Windsor, ON, Canada

Riad, Khaled Ain Shams University Cairo, Egypt

Ragueneau, Frederic ENS-Cachan Cachan, France

Rieder, Klaus WR Grace and Co-Conn Cambridge, MS, United States

Rahal, Khaldoun Kuwait University Safat, Kuwait

Rinaldi, Zila University of Rome Tor Vergata Rome, Italy

Rahman, Muhammad King Fahd University of Petroleum & Minerals Dhahran, Eastern Province, Saudi Arabia

Riva, Paolo University of Bergamo Dalmine, Italy

Rai, Durgesh Indian Institute of Technology Kanpur Kanpur, UP, India

Rivard, Patrice Université de Sherbrooke Sherbrooke, QC, Canada

Raj, Bharati Rajadhani Institute of Engineering and Technology Trivandrum, Kerala, India

Rivero-Angeles, Francisco Mexico, Distrito Federal, Mexico

Rakoczy, Anna University of Nebraska Lincoln, NE, United States Ramamurthy, K. IIT Madras Chennai, Tamilnadu, India Ramaswamy, Ananth Indian Institute of Science Bangalore, Karnataka, India Ramin, Keyvan Kermanshah, Islamic Republic of Iran Ramos, António Monte de Caparica, Portugal Ramyar, Kambiz Ege University Izmir, Turkey Randl, Norbert Carinthia University of Applied Sciences Spittal/Drau, Austria Rangan, Vijaya Curtin University of Technology Perth, Western Australia, Australia Rangaraju, Prasada Rao Clemson University Clemson, SC, United States Rao, Sarella National Institute of Technology Warangal, Andhra Pradesh, India Rasol, Mezgeen Dohuk Polytechnic University Zakho, Duhok, Iraq Ray, Indrajit Purdue University Calumet Hammond, IN, United States Reda Taha, Mahmoud University of New Mexico Albuquerque, NM, United States Regan, Paul Trigram London, UK

ACI Materials Journal/March-April 2016

Rizk, Emad Memorial University of Newfoundland St. John’s, NL, Canada Rizwan, Syed Ali University of Engineering and Technology Lahore, Punjab, Pakistan Rodrigues, Conrado Federal Centre for Technological Education in Minas Gerais, CEFET-MG Belo Horizonte, Minas Gerais, Brazil Rodríguez, Ángel Polytechnic University of Burgos Burgos, Spain Roh, Hwasung Chonbuk National University Jeonju, Jeollabuk-do, Republic of Korea Roshavelov, Theodore VSU Lyuben Karavelov Sofia, Bulgaria Rukzon, Sumrerng Rajamangala University of Technology Phra Nakhon Bangkok, Thailand Russo, Gaetano University of Udine Udine, Italy Saatci, Selcuk Izmir Institute of Technology Izmir, Turkey Sabet Divsholi, Bahador Nanyang Technological University Singapore Sadiq, Muhammad National University of Sciences and Technology Risalpur, Pakistan Sadowska-Buraczewska, Barbara Bialystok University of Technology Bialystok, Poland Saeki, Tatsuhiko Niigata University Niigata, Japan

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REVIEWERS IN 2015 Safan, Mohamed Menoufia University Shebeen El-Koom, Menoufia, Egypt

Sengul, Ozkan Istanbul Technical University Istanbul, Turkey

Safi, Brahim Research Unit Materials, Processes and Environment UR MPE/ Laboratory of Rheology/University of Boumerdes Boumerdes, Algeria

Sengupta, Amlan Indian Institute of Technology Madras Chennai, Tamil Nadu, India

Sagaseta, Juan University of Surrey Guildford, Surrey, UK Sahmaran, Mustafa Gazi University Ankara, Turkey Sajedi, Fathollah University of Malaya Kuala Lumpur, Selangor, Malaysia Sajedi, Siavash Ardabil, Islamic Republic of Iran Salem, Hamed Cairo University Giza, Egypt

Sengupta, Piyali National University of Singapore Singapore Seyed Kolbadi, Seyed Mahdi K.N. Toosi University of Technology Gorgan, Golestan, Islamic Republic of Iran Shabakhty, Naser University of Sistan and Baluchestan Zahedan, Sistan and Baluchestan, Islamic Republic of Iran Shafigh, Payam Kuala Lumpur, Malaysia Shafiq, Nasir University Technology Petronas Tronoh, Perak, Malaysia

Salib, Sameh Markham, ON, Canada

Shah, Santosh Dharmsinh Desai University Nadiad, Gujarat, India

Sancak, Emre Suleyman Demirel University Isparta, Turkey

Shahnewaz, Md University of British Columbia Kelowna, BC, Canada

Sánchez, Isidro University of Alicante Alicante, Spain

Shahzada, Khan KPK University of Engineering & Technology Peshawar, KPK, Pakistan

Sánchez, José Instituto Tecnológico de Oaxaca Oaxaca, Mexico

Shaikh, Fawad Stanford, CA, United States

Santos, Luis LNEC Lisboa, Portugal Saqan, Elias American University in Dubai Dubai, United Arab Emirates Sato, Ryoichi Hiroshima University Higashi-Hiroshima, Japan Sato, Yuichi Kyoto University Kyoto, Japan Scanlon, Andrew Pennsylvania State University University Park, PA, United States Schoepfer, Joan University of New Mexico Santa Fe, NM, United States Schwetz, Paulete Universidade Federal do Rio Grande do Sul Porto Alegre, Rio Grande do Sul, Brazil Sener, Siddik Istanbul Bilgi University Instanbul, Eyup, Turkey

260

Shao, Yixin McGill University Montreal, QC, Canada Sharifi, Yasser Vali-e-Asr University of Rafsanjan Rafsanjan, Islamic Republic of Iran Sharma, Akanshu Institute of Construction Materials Stuttgart, Germany Shawky, Mostafa Alexandria, Egypt Shayan, Ahmad ARRB Group Vermont South, Victoria, Australia She, Wei Southeast University NanJing, China Shehata, Medhat Ryerson University Toronto, ON, Canada Sheikh, Shamim University of Toronto Toronto, ON, Canada Shi, Haijian Kal Krishnan Consulting Services, Inc. Oakland, CA, United States

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Shi, Tao Zhejiang University of Technology Hangzhou, China

Soltanzadeh, Fatemeh ISISE Guimaraes, Minho, Portugal

Shi, Xianming Washington State University Pullman, WA, United States

Sonebi, Mohammed Queen’s University Belfast Belfast, UK

Shi, Xudong Tsinghua University Beijing, China

Song, Xuefeng Xi’an University of Architecture and Technology Xi’an, Shaanxi Province, China

Shi, Yilei Rockville, MD, United States

Sossou, Gnida Kwame Nkrumah University of Science and Technology  (KNUST) Kumasi, Ghana

Shivali, Ram Central Soil and Materials Research Station New Delhi, India Shuraim, Ahmed King Saud University Riyadh, Saudi Arabia Siad, Hocine Ryerson University Toronto, ON, Canada Silfwerbrand, Johan KTH Royal Institute of Technology Stockholm, Sweden Silva, Jessica University of Wisconsin–Madison Madison, WI, United States Singh, Harvinder Guru Nanak Dev Engineering College Ludhiana, Punjab, India Singh, Shamsher Birla Institute of Technology and Science, Pilani Campus Pilani, Rajasthan, India Siva Kumar, M. V. N. National Institute Of Technology-Warangal Warangal, Andhra Pradesh, India Smadi, Mohammad Jordan University of Science and Technology Irbid, Jordan Smith, Scott Southern Cross University Lismore, New South Wales, Australia Sneed, Lesley Missouri S&T Rolla, MO, United States So, Hyoung-Seok Seonam University Namwon, Republic of Korea Sobuz, Md. Habibur The University of Adelaide Adelaide, South Australia, Australia Soliman, Ahmed Western University London, ON, Canada Soltani, Masoud Tarbiat Modares University Tehran, Islamic Republic of Iran

ACI Materials Journal/March-April 2016

Souza, Regina Helena Universidade do Estado do Rio de Janeiro Rio de Janeiro, Brazil Söylev, Altug Yeditepe University Istanbul, Turkey Spadea, Giuseppe University of Calabria Cosenza, Italy Spainhour, Lisa FAMU-FSU Tallahassee, FL, United States Spinella, Nino University of Messina Messina, Italy Spyridis, Panagiotis Institute for Structural Engineering Vienna, Austria Stanton, John University of Washington Seattle, WA, United States Stivaros, Pericles Feld, Kaminetzky & Cohen, P.C. Jericho, NY, United States Su, Yu-Min National Kaohsiung University of Applied Sciences Sanmin, Taiwan, China Sujjavanich, Suvimol Kasetsart University Bangkok, Thailand Suksawang, Nakin Florida Institute of Technology Melbourne, FL, United States Sun, Zhihui University of Louisville Louisville, KY, United States Tadayon, Mohammad Hosein University of Tehran Tehran, Islamic Republic of Iran Tadros, Maher E.Construct.US, LLC Omaha, NE, United States Tae, Ghi Ho Leader Industrial Co. Seoul, Republic of Korea

261

REVIEWERS IN 2015 Takase, Yuya Tobishima Corporation Noda-shi, Chiba, Japan

Tepfers, Ralejs Ralejs Tepfers Consulting Göteborg, Sweden

Tan, Kefeng Southwest University of Science and Technology Sichuan, China

Thiagarajan, Ganesh University of Missouri–Kansas City Kansas City, MO, United States

Tan, Kiang Hwee National University of Singapore Singapore

Thomas, Adam Europoles GmbH Neumarkt, Germany

Tan, Sinjaya University of Houston Houston, TX, United States

Thomas, Blessen Thiruvalla, Kerala, India

Tang, Chao-Wei Cheng-Shiu University Niaosong District, Kaohsiung City, Taiwan, China Tang, Liqun South China University of Technology Guangzhou, Guangdong, China

Thorstensen, Rein Terje University of Agder Grimstad, Norway Tian, Ying University of Nevada, Las Vegas Las Vegas, NV, United States

Tang, Pei Eindhoven, the Netherlands

Tixier, Raphael Western Technologies Inc. Phoenix, AZ, United States

Tangtermsirikul, Somnuk Sirindhorn International Institute of Technology Patumthani, Thailand

Tolentino, Evandro Centro Federal de Educação Tecnológica de Minas Gerais Timóteo, Minas Gerais, Brazil

Tankut, Tugrul Middle East Technical University Ankara, Turkey

Topçu, İlker Eskişehir Osmangazi University Eskişehir, Turkey

Tapan, Mücip Yuzuncu Yil University Van, Turkey

Torrado-Gomez, Luz Universidad Pontificia Bolivariana Seccional Bucarmanga Bucaramanga, Santander, Colombia

Tarighat, Amir Tehran, Islamic Republic of Iran

Tosun, Kamile Dokuz Eylul University Izmir, Turkey

Tassios, Theodosios Athens, Greece Tastani, S. P. Demokritus University of Thrace Xanthi, Greece Tavio Sepuluh Nopember Institute of Technology (ITS) Surabaya, East Java, Indonesia Tawana, M. M. Tongji University Shanghai, China Tawfic, Yasser Minia University Minia, Egypt Taylor, Michael Granite Rock Co. Sacramento, CA, United States Tazarv, Mostafa University of Nevada, Reno Reno, NV, United States Tegos, Ioannis Salonica, Greece Teo, Wee Universiti Teknologi Petronas Tronoh, Perak Darul Ridzuan, Malaysia

262

Triantafillou, Thanasis University of Patras Patras, Greece Tsubaki, Tatsuya Yokohama National University Yokohama, Japan Tsuruta, Hiroaki Kansai University Suita, Japan Tuchscherer, Robin Northern Arizona University Flagstaff, AZ, United States Tuken, Ahmet King Saud University Riyadh, Saudi Arabia Turgut, Paki Harran University Sanliurfa, Turkey Tutikian, Bernardo Unisinos Porto Alegre, Rio Grande do Sul, Brazil Ueda, Naoshi Nagoya University Nagoya, Aichi, Japan Vaiciukyniene, Danute Kaunas, Lithuania

ACI Materials Journal/March-April 2016



REVIEWERS IN 2015

Van Deventer, J. S. J. University of Melbourne Melbourne, Victoria, Australia

Wardhono, Arie The State University of Surabaya Surabaya, Jawa Timur, Indonesia

Vaz Rodrigues, Rui EPFL Lausanne, VD, Switzerland

Watkins, Melanie Michigan Technological University Houghton, MI, United States

Vazquez-Herrero, Cristina Civil Engineering School La Coruña, Spain

Wei, Ya Tsinghua University Beijing, China

Veen, Cornelis Delft University of Technology Delft, the Netherlands

Wei-Jian, Yi Changsha, China

Velázquez Rodríguez, Sergio Universidad Panamericana Zapopan, Jalisco, Mexico Vellalapalayam Nallagounder, Vijayakumar Bannari Amman Institute of Technology Erode, Tamilnadu, India Venkatesh Babu, D. L. Kumaraguru College of Technology Coimbatore, Tamil Nadu, India Venkiteela, Giri NJDOT Trenton, NJ, United States Vercher, Jose Polytechnic University of Valencia Valencia, Spain Vimonsatit, Vanissorn Curtin University Perth, Western Australia, Australia Vintzileou, Elizabeth National Technical University of Athens Athens, Greece Viviani, Marco HEIG-VD Yverdon les Bains, Switzerland Vollum, Robert London, UK Wagh, Prabhanjan University of Cincinnati Cincinnati, OH, United States Waldron, Christopher University of Alabama at Birmingham Birmingham, AL, United States Wang, Huanzi San Jose, CA, United States Wang, Jingquan Southeast University Nanjing, Jiangsu, China Wang, Xiao-Yong Kangwon National University Chuncheon, the Republic of Korea Wang, Yuli Henan Polytechnic University Jiaozuo, Henan, China

ACI Materials Journal/March-April 2016

Weiss, Jason Purdue University West Lafayette, IN, United States Werner, Anne SIUE Edwardsville, IL, United States Wheat, Harovel University of Texas at Austin Austin, TX, United States Wheeler, Andrew University of Western Sydney Sydney, New South Wales, Australia Williams, Rupert University of the West Indies Saint Augustine, Trinidad and Tobago Wilson, William Universite de Sherbrooke Sherbrooke, QC, Canada Windisch, Andor Karlsfeld, Germany Wittmann, Folker Aedificat Institute Freiburg Freiburg, Germany Wong, Sook-Fun Nanyang Technological University Singapore Wood, Richard University of California, San Diego La Jolla, CA, United States Woyciechowski, Piotr Warsaw University of Technology Warsaw, Poland Wu, Chenglin Missouri S&T Rolla, MO, United States Wu, Hui Beijing, China Wu, Yu-You Dania Beach, FL, United States Xia, Jin Zhejiang University Hangzhou, Zhejiang, China Xiang, Tianyu Chengdu, Sichuan, China

263

REVIEWERS IN 2015 Xiangguo, Wu Harbin Institute of Technology Harbin, Heilongjiang, China

Yoon, Young-Soo Korea University Seoul, Republic of Korea

Xiao, Xiao Zhejiang University Hangzhou City, Zhejiang Province, China

Youzhi, Liu China Institute of Water Resources and Hydropower Research Beijing, China

Xiao, Yan Hunan University Changsha, Hunan, China

Yu, Baolin Michigan State University East Lansing, MI, United States

Xingyi, Zhu Hangzhou, China

Yüksel, Isa Bursa Technical University Bursa, Turkey

Xu, Aimin ARRB Group Melbourne, Victoria, Australia Xuan, D. X. The Hong Kong Polytechnic University Kowloon, Hong Kong, China Yakoub, Haisam Ottawa, ON, Canada Yaman, Ismail Middle East Technical University Ankara, Turkey Yang, Changhui College of Materials Science and Engineering Chongqing, China Yang, Keun-Hyeok Kyonggi University Suwon, Kyonggi-Do, Republic of Korea Yang, Yuguang Delft, the Netherlands Yang, Zhifu Middle Tennessee State University Murfreesboro, TN, United States Yazıcı, Şemsi Ege University Izmir, Turkey Yehia, Sherif American University of Sharjah Sharjah, United Arab Emirates Yekrangnia, Mohammad Sharif University of Technology Tehran, Islamic Republic of Iran Yepez, Fabricio Universidad San Francisco de Quito Quito, Ecuador Yerramala, Amarnath Dundee University Dundee, Scotland, UK Yildirim, Hakki Istanbul, Turkey Yilmaz, Bulent Bilecik Seyh Edebali University Bilecik, Turkey Yoon, Hyeong Jae Taisei Corporation Tokyo, Japan

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Zaidi, S. Kaleem Aligarh Muslim University Aligarh, UP, India Zatar, Wael West Virginia University Institute of Technology Montgomery, WV, United States Zayed, Abla University of South Florida Tampa, FL, United States Zeris, Christos National Technical University of Athens Zografou, Greece Zhang, Nan Nanjing Technical University Nanjing, Jiansu, China Zhang, Peng Karlsruhe Institute of Technology (KIT) Karlsruhe, Germany Zhang, Wei Ping Tongji University Shanghai, China Zhang, Y. X. The University of New South Wales Canberra, Australian Capital Territory, Australia Zhang, Yamei Southeast University Nanjing, China Zhao, Tiejun Qingdao, China Zhao, Xinyu State Key Laboratory of Subtropical Building Science Guangzhou, Guangdong, China Zheng, Herbert Gammon Construction Limited Hong Kong, China Zheng, Yulong Kyushu Institute of Technology Kitakyushu, Fukuoka, Japan Zhou, Changdong Beijing Jiaotong University Beijing, China Zhou, Shengjun MacGregor, Queensland, Australia Zhu, Han TianJin University TianJin, China

ACI Materials Journal/March-April 2016

ACI MATERIALS J O U R N A L J O U R N

The American Concrete Institute (ACI) is a leading authority and resource worldwide for the development and distribution of consensus-based standards and technical resources, educational programs, and certifications for individuals and organizations involved in concrete design, construction, and materials, who share a commitment to pursuing the best use of concrete. Individuals interested in the activities of ACI are encouraged to explore the ACI website for membership opportunities, committee activities, and a wide variety of concrete resources. As a volunteer member-driven organization, ACI invites partnerships and welcomes all concrete professionals who wish to be part of a respected, connected, social group that provides an opportunity for professional growth, networking, and enjoyment.